Statnamic Lateral Load Testing and Analysis of a Drilled Shaft in Liquefied Sand

Transcription

1 Brigham Young University BYU ScholarsArchive All Theses and Dissertations Statnamic Lateral Load Testing and Analysis of a Drilled Shaft in Liquefied Sand Seth I. Bowles Brigham Young University - Provo Follow this and additional works at: Part of the Civil and Environmental Engineering Commons BYU ScholarsArchive Citation Bowles, Seth I., "Statnamic Lateral Load Testing and Analysis of a Drilled Shaft in Liquefied Sand" (25). All Theses and Dissertations This Thesis is brought to you for free and open access by BYU ScholarsArchive. It has been accepted for inclusion in All Theses and Dissertations by an authorized administrator of BYU ScholarsArchive. For more information, please contact scholarsarchive@byu.edu, ellen_amatangelo@byu.edu.

2 STATMANIC LATERAL LOAD TESTING AND ANALYSIS OF A DRILLED SHAFT IN LIQUEFIED SAND by Seth I. Bowles A thesis submitted to the faculty of Brigham Young University in partial fulfillment of the requirements for the degree of Master of Science Department of Civil and Environmental Engineering Brigham Young University December 25

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4 Copyright 25 Seth Isaac Bowles All Rights Reserved

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6 BRIGHAM YOUNG UNIVERSITY GRADUATE COMMITTEE APPROVAL of a thesis submitted by Seth I. Bowles This thesis has been read by each member of the following graduate committee and by majority vote has been found to be satisfactory. Date Kyle M. Rollins, Chair Date Travis M. Gerber Date Steven E. Benzley

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8 BRIGHAM YOUNG UNIVERSITY As chair of the candidate s graduate committee, I have read the thesis of Seth I. Bowles in its final form and have found that (1) its format, citations, and bibliographical style are consistent and acceptable and fulfill university and department style requirements; (2) its illustrative materials including figures, tables, and charts are in place; and (3) the final manuscript is satisfactory to the graduate committee and is ready for submission to the university library. Date Kyle M. Rollins Chair, Graduate Committee Accepted for the Department E. James Nelson Graduate Coordinator Accepted for the College Alan R. Parkinson Dean, Ira A. Fulton College of Engineering and Technology

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10 ABSTRACT STATMANIC LATERAL LOAD TESTING AND ANALYSIS OF A DRILLED SHAFT IN LIQUEFIED SAND Seth Isaac Bowles Department of Civil and Environmental Engineering Master of Science Three progressively larger statnamic lateral load tests were performed on a 2.59 m diameter drilled shaft foundation after the surrounding soil was liquefied using downhole explosive charges. An attempt to develop p-y curves from strain data along the pile was made. Due to low quality and lack of strain data, p-y curves along the test shaft could not be reliably determined. Therefore, the statnamic load tests were analyzed using a ten degree-of-freedom model of the pile-soil system to determine the equivalent static load-deflection curve for each test. The equivalent static load-deflection curves had shapes very similar to that obtained from static load tests performed previously at the site. The computed damping ratio was 3%, which is within the range of values derived from the log decrement method. The computer program LPILE was then used to compute the load-deflection curves in comparison with the response from the field load tests. Analyses were

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12 performed using a variety of p-y curve shapes proposed for liquefied sand. The best agreement was obtained using the concave upward curve shapes proposed by Rollins et al. (25) with a p-multiplier of approximately 8 to account for the increased pile diameter. P-y curves based on the undrained strength approach and the p-multiplier approach with values of.1 to.3 did not match the measured load-deflection curve over the full range of deflections. These approaches typically overestimated resistance at small deflections and underestimated the resistance at large deflections indicating that the p-y curve shapes were inappropriate. When the liquefied sand was assumed to have no resistance, the computed deflection significantly overestimated the deflections from the field tests.

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14 ACKNOWLEDGMENTS Dr. Rollins has been a wonderful professor to work with. He has been very patient with me and the difficulties we have had with the analysis. He has always been willing to help me when ever I needed it. He has had to spend many hours in my office helping me try to troubleshoot the analysis. Dr. Rollins patience with me and explaining difficult subjects has been invaluable. I also need to thank Dr. Gerber for all of his help and the use of PY_BYU for deriving the p-y curves from my strain data. Almost an entire summer was spent here in room 192 of the Clyde Building helping me figure out how to use his program. He was also very willing to answer any questions that I had and if he didn t know the answer right away he would find it. My wonderful and supportive wife Aimee deserves an award. She has supported me through the majority of my schooling here at BYU. She has put up with all the late nights and boring topics I have come home talking about. Now she is also bearing the tiring task of caring for of our new baby girl, Madison, who has her sleep schedule mixed up. Without her love and support, I do not think I would have been able to make it through the last bit of my schooling.

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16 The funding for this research was provided by the National Science Foundation (NSF) under Grant No. CMS The support was very appreciated. The views and recommendations expressed in this thesis are not necessarily the views of NSF.

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18 TABLE OF CONTENTS LIST OF TABLES... xiii LIST OF FIGURES...xv 1 Introduction and Objectives Introduction Objective and Scope of Research Background Regarding P-Y Curves and Their Development Background Introduction P-Y Curve Information Prior to Full-Scale Testing TILT Project P-Y Curves Developed From TILT Testing Estimating P-Multiplier Adjustments for Diameter Static Test on Drilled Shaft MP-1 at the Mt. Pleasant Site Brief History of the Statnamic Device (Bermingham 2) Statnamic Test on Drilled Shaft MP-3 at the Mt. Pleasant Site Current Research Focus Site and Soil Description Site Location and Bridge Description Geological Background Scope of Geotechnical Investigation...33 ixix

19 3.4 Test Borings and Laboratory Investigations In-Situ Testing Liquefaction Hazard Analysis Test Set-Up and Pile Description Introduction Pile Description Test Set-Up Above Ground Instrumentation Below Ground Instrumentation Blast Layout Statnamic Lateral Load Test Results Introduction Lateral Load Tests Pile Motion from Acceleration Data Piezometer Data Comparison of the Three Load Test Results Analysis Introduction Calculating P-Y Curves from Strain Data Empirical Evaluation Conclusions Introduction Blast Induced Liquefaction Statnamic Versus Earthquake xx

20 7.4 Static Versus Statnamic Stiffness Dynamic Versus Static Loads Static Load Deflection Curves Concave Up P-Y Curves Lateral Resistance in Liquefied Sand Analysis Versus Existing Methods Recommendations References Appendix A Additional Information from Testing xixi

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22 LIST OF TABLES Table 5-1 Comparison of rise time, lag time, peak acceleration, and peak velocity Table 5-2 Ru values after detonation, before and after loading for load test Table 5-3 Ru values after detonation, before and after loading for load test Table 5-4 Ru values after detonation, before and after loading for load test Table 6-1 Linear stiffness, natural period, and damping ratio used for each test Table 6-2 Soil properties used in the analysis of Rollins et al., (25) comparison Table 6-3 Soil properties used in the analysis and comparison to the Matlock (197) and Wang and Reese (1998) model Table 6-4 Soil properties used in the analysis and comparison to the Liu and Dobry (1995) and Wilson (1998) p-multiplier models Table A-1 Settlement of Mt. Pleasant test site Table A-2 Compressive strengths of concrete used for the construction of MP xiii

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24 LIST OF FIGURES Figure 1-1 Derivation process used to develop p-y curves from strain measurements (Hales, 23)...6 Figure 2-1 Pile head lateral load versus displacement curves for 324 mm steel pipe piles before and after liquefaction based on Treasure Island liquefaction testing program (Ashford and Rollins, 2) Figure 2-2 Pile head load and excess pore pressure ratio as a function of time for single pipe pile test at Treasure Island (written communication Kyle Rollins)...12 Figure 2-3 Measured load-displacement curves for a single pile in non-liquefied and liquefied sand in comparison with curves computed using several values of residual strength (written communication, Kyle Rollins) Figure 2-4 Expected shape of p-y curve for liquefied sand in contrast to soft clay curve shape (written communication, Kyle Rollins)...14 Figure 2-5 Summary of calculated p-y curves for the east center pile in the 3x3 pile group during the first post-blast load series (Gerber, 23) Figure 2-6 Summary of calculated p-y curves for the east center pile in the 3x3 pile group during the tenth post-blast load series (Gerber, 23)...18 Figure 2-7 Post-blast p-y curves for the east center pile of the 3x3 pile group at various depths during the first and tenth load series, where average R u is shown as a percent (Gerber, 23) xv

25 Figure 2-8 Post-blast p-y curves for the east center pile of the 3x3 pile group during the first load series (Gerber, 23)...2 Figure 2-9 Post-blast p-y curves for the east center pile of the 3x3 pile group during the tenth load series (Gerber, 23)....2 Figure 2-1 Applied load versus pile head displacement curves for all three load tests from testing on MP-1 (Hales, 23) Figure 2-11 Peak pore pressure ratio with depth immediately after the first blast for MP-1, at the Mt. Pleasant test site...24 Figure 2-12 Peak pore pressure ratio with depth immediately after the second blast for MP-1, at the Mt. Pleasant test site Figure 3-1 Aerial photograph of Cooper River bridges and test site. (taken from a presentation by the SCDOT and S&ME to the CE Club)...3 Figure 3-2 Site map showing approximate locations of SPT and CPT borings (a) relative to the existing bridge approach ramps and (b) relative to the test site (Brown, 2)...31 Figure 3-3 Artist's rendering of future Ravenel Bridge (Bridgepros, 25)...32 Figure 3-4 Boring log for test hole DS-1 (Hales 23) Figure 3-5 Boring log for test hole MPS-11 (Hales 23) Figure 3-6 Boring log for test hole LB-28 (Hales 23) Figure 3-7 Idealized soil profile for the Mt. Pleasant test site (Modified from Camp et al., 2a) Figure 3-8 Atterberg limits tests at various depths within the Cooper River Marl relative to the plasticity chart (Camp et al., 2b)...5 xvi

26 Figure 3-9 Natural moisture content versus elevation in the Cooper River Marl (Camp et al., 2b)...5 Figure 3-1 Fines content versus elevation in the Cooper River Marl (Camp et al., 2b) Figure 3-11 Undrained shear strength versus elevation from UU and CU triaxial shear test on undisturbed samples of the Cooper River Marl (Camp et al., 2b) Figure 3-12 Results of drained triaxial shear strength tests on Cooper River Marl plotted in a p-q diagram (Camp et al., 2b) Figure 3-13 Normalized SPT clean sand penetration resistance versus depth for three test holes near the test site...54 Figure 3-14 Interpreted relative density versus depth based on SPT penetration resistance for three holes close to the test site Figure 3-15 Results from CPT sounding LTB-1 including normalized cone resistance, friction ratio, and pore pressure along with interpreted relative density and soil profile...57 Figure 3-16 Results from CPT sounding MPS-7 including normalized cone resistance, friction ratio, and pore pressure along with interpreted relative density and soil profile...58 Figure 3-17 Results from CPT sounding GT-1 including normalized cone resistance friction ratio, and pore pressure along with interpreted relative density and soil profile...59 xvii

27 Figure 3-18 Interpreted Relative density and friction angle versus depth for sand layers in the soil profile based on three CPT soundings...62 Figure 3-19 Interpreted Undrained shear strength versus depth for clay layers in the soil profile based on three CPT soundings...63 Figure 3-2 Profiles of V s and V s1 versus depth based on down SCPT sounding and a down-hole shear wave velocity test conducted by Redpath Geophysics Figure 3-21 Photograph of a brick house wrecked by the Charleston earthquake of August 31, 1886 (USGS, 25) Figure 3-22 Photograph of a sand boil due to liquefaction during the 1886 Charleston, South Carolina Earthquake (FHWA, 25) Figure 3-23 Profiles showing cone tip resistance, SBT index, and factor of safety against liquefaction versus depth for GT-1 due to M7.3 earthquake producing.77 g peak acceleration associated with a 2% probability of exceedance in 5 years. (Hales, 23)...7 Figure 3-24 Profiles showing cone tip resistance, SBT index, and factor of safety against liquefaction versus depth for GT-1 due to M6.4 earthquake producing.16 g peak acceleration associated with a 1% probability of exceedance in 5 years. (Hales 23)...71 Figure 4-1 Contractor used a track-mounted SoilMec for drilling (photograph from a presentation by the SCDOT and S&ME to the CE Club) Figure 4-2 Photograph of worker assembling reinforcement cage at the Mount Pleasant site (photograph from a presentation by the SCDOT and S&ME to the CE Club) xviii

28 Figure 4-3 Drilled shaft dimensions, strain gauges, and accelerometers...76 Figure 4-5 Drilled shaft and corresponding soil profile...77 Figure 4-6 Schematic of statnamic load test at the Mt. Pleasant site (drawing modified from the Ravenel Bridge Project Load Test Plans)...78 Figure 4-7 Schematic of the statnamic loading device with accelerometers and LVDTs at the Mount Pleasant site (Figure provided by AFT Inc. load report for MP-3)....8 Figure 4-8 Reinforcement cage after installation of strain gages and inclinometers (photograph from a presentation by the SCDOT and S&ME to the CE Club)...82 Figure 4-9 Plan view of the piezometers and charges (Brown, 2)...83 Figure 4-1 Elevation view showing a profile of piezometers and down-hole charges relative to the test shaft...84 Figure 5-1 Pile head deflection time history for the first lateral load test on test pile MP Figure 5-2 Load time history for the first lateral load test on test pile MP Figure 5-3 Load versus deflection curve for load test 1 on test pile MP Figure 5-4 Pile head deflection time history for the second lateral load test on test pile MP Figure 5-5 Load time history for the second lateral load test on test pile MP Figure 5-6 Load versus Deflection for load test Figure 5-7 Pile head deflection time history for the third lateral load test on test pile MP Figure 5-8 Load time history for the third lateral load test on test pile MP xix

29 Figure 5-9 Load versus Deflection curve for load test Figure 5-1 Acceleration, velocity, and deflection graphs from test 1 accelerometers...97 Figure 5-11 (Continued) Acceleration, velocity, and deflection graphs from test 1 accelerometers...98 Figure 5-12 (Continued) Acceleration, velocity, and deflection graphs from load test 1 accelerometers...99 Figure 5-13 (Continued) Acceleration, velocity, and deflection graphs from load test 1 accelerometers...1 Figure 5-14 Acceleration versus depth plots plot at several times for load test Figure 5-15 Velocity versus depth plots derived from accelerations at several times for load test Figure 5-16 Deflection versus depth plots derived from accelerations at several times for load test 1 along with measured deflections from LVDTs above ground Figure 5-17 Acceleration, velocity, and deflection graphs from load test 2 accelerometers...13 Figure 5-18 Acceleration versus Depth time step plot from load test Figure 5-19 Velocity versus Depth time step plot from load test Figure 5-2 Deflection versus Depth time step plot from load test Figure 5-21 Acceleration, velocity, and deflection graphs from load test 3 accelerometers...19 Figure 5-22 Acceleration versus Depth time step plot from load test xx

30 Figure 5-23 Velocity versus Depth time step plot from load test Figure 5-24 Deflection versus Depth time step plot from load test Figure 5-25 R u time histories from the first blast for (a) B5, (b) B6, and (c) B Figure 5-26 R u time histories from the first blast for (a) B7, (b) B2, and (c) B Figure 5-27 R u time histories from the first blast for (a) B1, (b) B11, and (c) B Figure 5-28 R u time histories from the first blast for (a) B12, (b) B13, and (c) B Figure 5-29 Peak R u versus depth plots for the first load test immediately after the charges were detonated Figure 5-3 Peak R u versus depth for the first load test just after the statnamic device was fired Figure 5-31 R u time histories from the second blast for (a) B5, (b) B6, and (c) B Figure 5-32 R u time histories from the second blast for (a) B7, (b) B2, and (c) B Figure 5-33 R u time histories from the second blast for (a) B1, (b) B11, and (c) B Figure 5-34 R u time histories from the second blast for (a) B12, (b) B13, and (c) B Figure 5-35 Peak R u versus depth plots for the second load test immediately after the charges were detonated Figure 5-36 Peak R u versus depth plots for the second load test immediately after the statnamic device was fired Figure 5-37 R u time histories from the third blast for (a) B5, (b) B6, and (c) B Figure 5-38 R u time histories from the third blast for (a) B7, (b) B2, and (c) B xxi

31 Figure 5-39 R u time histories from the third blast for (a) B1, (b) B11, and (c) B Figure 5-4 R u time histories from the third blast for (a) B12, (b) B13, and (c) B Figure 5-41 Peak R u versus depth plots for the third load test immediately after the charges were detonated Figure 5-42 Peak R u versus depth plots for the third load test immediately after the statnamic device was fired Figure 5-43 Comparison of the three applied pile head load versus deflection curves Figure 5-44 Maximum positive and negative acceleration, velocity, and deflection for all three tests Figure 5-45 Comparison of the piezometer readings for the three tests Figure 5-46 Excess pore pressure ratio contours (in percent) for the soil profile mass immediately after the detonation of the charges for test Figure 5-47 Excess pore pressure ratio contours ( in percent) for the soil mass immediately after the statnamic loading for test Figure 5-48 Excess pore pressure ratio contours (in percent) for the soil mass immediately after the detonation of the charges for test Figure 5-49 Excess pore pressure ratio contours (in percent) for the soil mass immediately after the statnamic loading for test Figure 5-5 Excess pore pressure ratio contours (in percent) for the soil mass immediately after the detonation of the charges for test xxii

32 Figure 5-51 Excess pore pressure ratio contours (in percent) for the soil mass immediately after the statnamic loading for test Figure 6-1 Time step curvatures calculated from strain gauges for test 1 of MP Figure 6-2 Time step curvatures calculated from strain gauges for test 2 of MP Figure 6-3 Time step curvatures calculated from strain gauges for test 3 of MP Figure 6-4 Model used to calculate the inertial force (relative size of the masses provides an approximate indication of mass distribution) Figure 6-5 Deflection profile for load test 3 used to find the active length Figure 6-6 Comparison of load-deflection curves for test Figure 6-7 Comparison of load-deflection curves for test Figure 6-8 Comparison of load-deflection curves for test Figure 6-9 Plots of the measured statnamic force time history, computed inertia, damping and spring force time histories for test Figure 6-1 Plots of the measured statnamic force time history, computed inertia, damping and spring force time histories for test Figure 6-11 Plots of the measured statnamic force time history, computed inertia, damping and spring force time histories for test Figure 6-12 Comparison of the static equivalent load-deflection curves for all three tests Figure 6-13 Average pore water pressures for the first blast, first cycle and the second blast, first cycle compared to the average pore pressures of all three load tests of MP xxiii

33 Figure 6-14 Comparison of the load deflection curve of the first blast of MP-1 and the static equivalent load deflection curve of MP Figure 6-15 Comparison of the load deflection curve of the second blast of MP-1 and the static equivalent load deflection curve of MP Figure 6-16 Relationship between residual strength and corrected SPT resistance (Seed and Harder, 199) Figure 6-17 Comparison of the use of soft clay p-y curve for liquefied sand versus the calculated static equivalent for test 2 of MP Figure 6-18 Comparison of the method used by Liu and Dobry (1995) and Wilson (1998) compared to the calculated equivalent static stiffness of MP Figure A-1 MP-3 drilled shaft alignment measured by Trevi Icos Corporation Figure A-2 Graph showing the recorded settlement for the Mt. Pleasant test site while testing MP xxiv

34 1 Introduction and Objectives 1.1 Introduction The lateral load capacity of deep foundations is critically important in the design of bridges, buildings and other structures in seismically active regions. Although fairly reliable methods have been developed for predicting the lateral resistance of piles in nonliquefied soils, there is little information to guide engineers in the design of piles that are surrounded by liquefiable soils. Without an accurate assessment of the resistancedisplacement relationship for piles in liquefied soils, it becomes impossible to determine whether additional piles may be necessary for a foundation in liquefied sand or whether soil improvement must be undertaken to inhibit the development of liquefaction. Improper assessments can lead to seismically unsafe structures or unnecessary expense. These issues become even more important as the engineering profession attempts to move to performance-based design codes where estimates of displacements are required. While ongoing centrifuge studies using small-scale models can provide valuable insights, full-scale tests are necessary to verify/calibrate these models and provide ground truth information. 1

35 1.2 Objective and Scope of Research The world s first full-scale lateral pile load tests, utilizing controlled blasting to achieve liquefaction within the surrounding soil, were performed during 1998 and 1999 at Treasure Island in San Francisco Bay. This thesis describes the second set of full-scale laterally loaded tests involving blast-induced liquefaction which were conducted near Charleston, South Carolina in 2. The testing in Charleston provides a valuable opportunity to expand and supplement the data and results that were obtained from Treasure Island. For example, the diameters of the pile foundations at the Treasure Island test site were typically about one eighth the diameter of the shaft foundations used in Charleston. This difference allows for an evaluation of the effect of a much wider and stiffer pile on the p-y cures. In addition, the liquefied thickness at Treasure Island was on the order of 6 to 8 m while that at the Charleston site was about 12 m. This deeper liquefied zone makes it possible to evaluate the effect of greater initial effective stress (or greater depth) on the p-y curves. Finally, in contrast to the Treasure Island tests, which were conducted statically using only hydraulic actuators, the tests in Charleston were conducted both statically and dynamically using a statnamic rocket sled to apply load in about.2 seconds. These test results make it possible to evaluate the influence of rate of loading and damping on the measured lateral resistance and p-y curves. The analysis of the static testing at Charleston was the subject of a thesis by Hales (23), while this thesis will focus on the analysis of the dynamic testing. 2

36 The overall objective of this study is to better understand the resistance the large diameter deep foundations provide in liquefied soil through full-scale testing. Specifically, the objectives of this study are to: 1. Develop p-y curves for the liquefied soil to determine whether they correlate with the concave-up shaped curves developed by Wilson (1998) through centrifuge testing and those resulting from analysis of the Treasure Island Tests (Ashford and Rollins, 22; Rollins et al., 25). 2. Evaluate the influence of increasing depth, initial effective stress and excess pore pressure ratio on p-y curves in liquefied sand. 3. Quantify the effect of a stiffer, larger diameter pile on the generated p-y curves when compared with those derived by Rollins et al., (25) for smaller diameter piles. 4. Determine the effect of dynamic loading on the lateral resistance of a pile in liquefied sand relative to the static resistance. The South Carolina Department of Transportation (SCDOT) provided funding for the conventional axial and lateral load tests as well as the liquefaction load tests. The tests were performed on the Mt. Pleasant side of the Cooper River near the location where the Ravenel Bridge was proposed to be built. Modern Continental South, Inc. served as the general contractor for the testing project and supervised the Mt. Pleasant site testing. The test results were originally intended to aid in the design of the bridge, but Dr. Rollins of Brigham Young University was able to procure a grant from the National Science Foundation to allow for a more detailed analysis of the data. Therefore, this study benefits from $25, already spent by the SCDOT for the foundation testing 3

37 and instrumentation. Although there were many tests performed at the Mt. Pleasant site, this thesis will only focus on the analysis and interpretation of the data collected from the statnamic lateral load test in liquefied soil performed on the foundation labeled MP Background Regarding P-Y Curves and Their Development The lateral resistance of a deep foundation is a function of both the structural stiffness of the foundation itself and the resistance of the surrounding soil. Therefore, engineering analysis of soil-structure interaction problems such as this requires accurate assessments of the non-linear behavior of both the surrounding soil and the foundation. The lateral resistance which the soil provides is a non-linear function of the lateral deflection of the foundation. A graphical representation of the relationship between resistance and deflection is portrayed graphically as a p-y curve. The p-y curve is plotted with the horizontal deflection (y) on the abscissa or the x-axis and the soil resistance expressed as a force per length of foundation (p) on the ordinate or the y-axis. Soil type plays a significant role in the variation of the stiffness and shape of p-y curves. Other important factors include pile diameter, embedment depth, and various soil properties such as strength and unit weight. In general, the lateral resistance (p) tends to increase with increasing diameter of pile and with increasing depth below the ground surface. A number of investigators have developed equations for p-y curves in stiff clay, soft clay and non-liquefied sand; however, considerable uncertainty exists regarding appropriate p-y curves for liquefied sand and how these curves might be affected by initial vertical stress and pile diameter. Insight into factors to account for these effects can be obtained from full-scale testing. 4

38 Because it is impossible to directly measure the deflection and soil pressure with depth, an indirect method has been used along with basic beam theory. Strain gauges along the length of the pile allow for the curvature of the pile to be evaluated. From this curvature the pile deflection and soil pressure can be calculated. The first assumption needed to derive the deflections is that the pile acts like an idealized Timoshenko beam. This means that deflections are a result of bending only, and deflections due to shear are neglected. These assumptions are only valid in long slender beams. Most piles fit the criteria of being a slender beam. Deflection is calculated from double integration of the curvatures with respect to the pile length. To be able to do this integration, a cantilever support condition is often assumed, where the deflection and curvature at the bottom end of the pile is assumed to be zero. Once again, this assumption is generally acceptable for deep foundations. Moment can be derived from curvature by multiplying the curvature by the appropriate bending stiffness (or EI). From the moment, the pressure can be derived through double differentiating moment with respect to distance along the pile. Since pressure derived from moment is a material dependent calculation, the non-linear EI for a concrete pile must be accurately estimated to reliably compute pressure. This relationship will be discussed further in Chapter 6. Figure 1-1 gives a step by step process used to develop p-y curves from strain data. 5

39 Figure 1-1 Derivation process used to develop p-y curves from strain measurements (Hales, 23). 6

40 2 Background 2.1 Introduction Centrifuge model testing has been the primary method for evaluating the lateral resistance of piles in liquefied soils. Although model testing is important because it facilitates parametric studies, it can t represent a full-scale test completely. Cost is the main reason why scale model testing is used and will continue to be used. Full-scale tests have been performed which allow us to compare the model test results with actual performance data representing a few parameter. Since full-scale testing provides the actual response of a foundation we can substantiate the results of model testing and then apply the combined results of both model and full-scale tests to foundation design with confidence. In Section 2.2, information regarding p-y curves for liquefied soils prior to full scale testing will be presented. Since the Treasure Island Liquefaction Test (TILT) program was the first full-scale test of its kind to be performed, a detailed review of this test will be given. Section 2.3 will review the preliminary test results from the TILT project. Section will review the p-y curves developed by Gerber (23) from the TILT project data and subsequently reported by Rollins et al, (25). Section 2.5 will discuss p-y curves as a function of pile diameter based on the TILT test results. Section 2.6 will introduce the lateral load testing program conducted at the M. Pleasant site near 7

41 Charleston, South Carolina, with a focus on the analysis of liquefied soil response under static loading. Section 2.7 will provide a brief history of statnamic testing. Section 2.8 will focus on the testing of liquefied soils at the Mt. Pleasant site using the statnamic device and the subsequent analysis of the soil response made by Brown (2) of ATF. Finally Section 2.9 will address the particular focus of this thesis. 2.2 P-Y Curve Information Prior to Full-Scale Testing Existing information regarding p-y curves for liquefied sand is still indefinite even though research in this field has been going on for some time. In 1995, a greater interest in this topic was initiated, and since then much more research has been conducted to solidify the opinions and research results to converge on a design methodology. Wang and Reese (1998) proposed that the resistance in liquefied sand can be explained by the p-y curve for soft-clay (e.g. Matlock, 197) if the ultimate strength is set equal to the undrained residual shear strength of sand. Wang and Reese use the work of Seed and Harder (199) to suggest the undrained residual shear strength of sand can be estimated using correlations with apparent relative density. Through centrifuge model testing in medium dense sand with a relative density of 6%, Liu and Dobry (1995) found that the ultimate strength of fully liquefied sand was one tenth its non-liquefied strength. So using this multiplier of.1 with p-y curves back calculated from tests in non-liquefied sand, a reasonable match was made with measured bending moments from a model pile. After more centrifuge tests in sand with a relative density of about 4%, Abdoun (1997) agreed with the.1 multiplier from Liu and Dobry. In other research efforts, Tokimatsu (1999) found that a p-multiplier ranging from.5 to 8

42 .2 gave good representations of the observed field performance of piles subject to lateral spreading. In other centrifuge studies conducted at U.C. Davis Wilson (1998) derived p-y curves for liquefied sand using a set of ground shaking time histories. Wilson compared his p-y curves to API (1993) sand and found the p-multiplier to be.1 to.2 for loose sand (~35% relative density) during peak loading cycles while the sand was liquefied. He also found that for medium dense sand (~55% relative density) that the p-multiplier was around.25 to.35. At different times in the loading time history p-multipliers of more that 1 existed and later on in the loading time history p-multipliers ranged from.1 to.35 after the soil had lost significant amount of resistance due to cyclic loading. Goh (21) is another person that used centrifuge testing to produce p-y curves. He used data from the results of Abdoun (1997) and analytical studies to try and develop a dimensionless p-y curve for liquefied sand. His resulting p-y curve shape was different from all the other researchers that used the p-multiplier approach. From the review of past research on p-y curves for liquefied soil, there is a need for further research like the TILT project and the current project in Charleston. Full-scale testing will hopefully shed a little more light on the soil resistance in liquefied soil. 2.3 TILT Project To improve our understanding of the lateral load behavior of deep foundations in liquefied soil, a series of lateral load tests were recently conducted on full-scale piles, pile groups, and drilled shaft foundations (Ashford and Rollins, 2; Ashford and Rollins, 22). The testing was conducted at the National Geotechnical Test Site on Treasure 9

43 Island in San Francisco Bay and is known as the Treasure Island Liquefaction Test, or TILT, program. Tests were performed after a surface layer of soil was liquefied using controlled blasting techniques. These tests were very successful and demonstrated that controlled blasting can induce liquefaction in a well-defined volume of soil in the field for full-scale experimentation. Excess pore pressure ratios (R u ) of 9 to 1% were generated within a depth range of 1 to at least 6 m and over a 13 m x 19 m surface area. R u values greater than 8% were typically maintained for 6 minutes (Rollins et al. 2). The TILT project represents the first full-scale tests ever performed on deep foundations in liquefied soils. A typical plot of load versus displacement for a lateral load test on a single pile at Treasure Island is shown in Figure 2-1. The test was performed using displacementcontrol procedures and forces applied to the pile head were measured with load cells. Initially, single cycles with maximum displacements of 75, 15, and 225 mm were applied, and then nine additional cycles were applied with a maximum displacement of 225 mm. As cycling continued, the load-displacement curves rapidly degraded to an S- shaped curve that was relatively consistent. A comparison with the load-displacement curve prior to liquefaction indicates that the reduction in strength following liquefaction is substantial. For the S-shape curve, very little resistance was developed initially, but after a displacement of about 75 mm there was a rapid increase in resistance with continued displacement. This increase in the pile-soil system stiffness appears to be tied to the development of reduced pore water pressures due to dilation of the sand following continued displacement. A time history of the measured excess pore pressure ratio (R u ) is shown in Figure 2-2 along with a time history of measured load on the pile. At the 1

44 beginning of each cycle R u is near 1%, indicating complete liquefaction. As displacement increases in each cycle, R u drops substantially and this drop in pore pressure produces a corresponding increase in lateral resistance Non-Liquefied Liquefied Load (kn) Displacement (mm) Figure 2-1 Pile head lateral load versus displacement curves for 324 mm steel pipe piles before and after liquefaction based on Treasure Island liquefaction testing program (Ashford and Rollins, 2). 11

45 Ru (%) Load (kn) Time (sec) Time (sec) Figure 2-2 Pile head load and excess pore pressure ratio as a function of time for single pipe pile test at Treasure Island (written communication Kyle Rollins). Preliminary lateral load analyses were performed to provide a rough assessment of existing methods for developing p-y curves in liquefied sand. Figure 2-3 shows the measured load-displacement relationships for one cycle before and after liquefaction. Load-displacement curves computed using the computer program LPILE (24) and are 12

46 also shown in Figure 2-3 for three different cases. IN two of the LPILE analyses, the liquefied sand was assumed to have a soft-clay p-y curve with a residual undrained strength equal to the lower-bound and average values obtained from correlation with the (N 1 ) 6 value for the sand using the relationship of Seed and Harder (199). At lower displacement levels, the computed resistance was significantly higher than the measured resistance, but at higher displacements, the measured resistance was within the range of computed values Load (kn) Displacement (mm) Non-Liquefied Liquefied LPILE-Avg Residual LPILE-Low Residual LPILE-No Residual Figure 2-3 Measured load-displacement curves for a single pile in non-liquefied and liquefied sand in comparison with curves computed using several values of residual strength (written communication, Kyle Rollins). In the third LPILE analysis, the liquefied sand was assumed to have no resistance at all. In this case, the computed load-displacement curve was very close to the measured curve at low displacements suggesting that there is little to no soil resistance acting. 13

47 However, at displacements greater than about 75 mm, the measured resistance rapidly increased beyond the computed value. These preliminary results suggested that the p-y curve for liquefied sand would have a shape that is concave upward as shown in Figure 2-4 which is in stark contrast with the soft clay curve shape that is typically assumed. However a concave down p-y curve shape similar to the soft clay curve was reported by Wilson (1998) in his interpretation of centrifuge tests. Liquefied Sand Based on Soft Clay Curve Horizontal Resistance/Length, P Liquefied Sand Suggested by Treasure Island Testing Horizontal Displacement, y Figure 2-4 Expected shape of p-y curve for liquefied sand in contrast to soft clay curve shape (written communication, Kyle Rollins). 14

48 2.4 P-Y Curves Developed From TILT Testing Results from the TILT project p-y curves are published in Gerber s dissertation, P-Y Curves for Liquefied Sand Subject to Cyclic Loading Based on Full-Scale Testing of Deep Foundations (23). A portion of the curves he developed will be subsequently presented. The results from the TILT project confirm the assumption that the p-y curves would present themselves as concave-up, reflective of the load-displacement curve for the pile head. Gerber analyzed a 3x3 group of 324 mm diameter steel pipe piles along with a single 324mm diameter steel pipe pile. In Gerber s dissertation, he presented p-y curves for five of the nine piles tested and the single pipe pile. After review of the results, the east pile in the center row appeared to be the most reliable and representative. Figure 2-5 shows the plots generated of the p-y relationship from the first load series of the tests for the first 1 strain gage depths or stations, and Figure 2-6 shows the p-y relationship from the tenth load series for the first 1 stations. Figure 2-7 shows comparisons of simplified, lower bound p-y curves from the first and tenth load series for the first 7 stations. Figure 2-8 and Figure 2-9 show the same p-y curves as displayed in Figure 2-7, but with the various curves for the same series on the same plot. From the TILT test, Gerber s analytical results, and study of p-y plots like the ones shown, the following conclusions are reached: 1. P-y curves for liquefied soils are characterized by a concave-up shape where the stiffness of the curve increases with displacement (as seen in Figure 2-7 through Figure 2-9). 15

49 2. The concave-up shape seems to result primarily from dilation of the soil due to shearing as the pile is displaced. Gapping effects, however, likely also contribute to the observed shape of the p-y curves. 3. The stiffness of p-y curves for liquefied soils increase with increasing depth (as seen in Figure 2-8 and Figure 2-9) and decreasing excess pore water pressure (as seen in Figure 2-7). The p-y curves appear to transition from a concave-up to a concave-down shape with decreasing excess pore water pressures. 4. As already mentioned in section 2.3, the concave-up shape of the p-y curves derived for liquefied sand is starkly different from the shape given from a residual undrained shear strength design approach. The same difference exists with a p-multiplier design approach (Gerber 23). Using LPILE, it was confirmed that the derived p-y curves yield better matches with the measured pile head deflections and moment curves over a large range of applied loads than the p-y curves using these two design approaches. 16

50 Load Point Depth Below Ground Surface (meters) 1 kn/m 2 cm 1 kn/m 2 cm. A A Avg. Ru = 94% B Avg. Ru = 76%.76 B 1.52 C 1 kn/m 1 kn/m 2.29 D 2 cm 2 cm 3.5 E C Avg. Ru = 85% D Avg. Ru = 98% 3.81 F 1 kn/m 1 kn/m 4.57 G 5.33 H E 2 cm 2 cm Avg. Ru = 1% F Avg. Ru = 1% 6.1 I 6.86 J 1 kn/m 1 kn/m cm 2 cm 8.38 G Avg. Ru = 98% H kn/m 1 kn/m cm 2 cm I J Figure 2-5 Summary of calculated p-y curves for the east center pile in the 3x3 pile group during the first post-blast load series (Gerber, 23). 17

51 Load Point Depth Below Ground Surface (meters) 1 kn/m 2 cm 1 kn/m 2 cm. A A Avg. Ru = 63% B Avg. Ru = 56%.76 B 1.52 C 1 kn/m 1 kn/m 2.29 D 2 cm 2 cm 3.5 E C Avg. Ru = 7% D Avg. Ru = 77% 3.81 F 1 kn/m 1 kn/m 4.57 G 5.33 H E 2 cm 2 cm Avg. Ru = 65% F Avg. Ru = 66% 6.1 I 6.86 J 1 kn/m 1 kn/m cm 2 cm 8.38 G Avg. Ru = 51% H kn/m 1 kn/m cm 2 cm I J Figure 2-6 Summary of calculated p-y curves for the east center pile in the 3x3 pile group during the tenth post-blast load series (Gerber, 23). 18

52 Load Point Depth Below Ground Surface (meters) 75 kn/m 25 kn/m kn/m 25 kn/m A A 5 cm 1 cm 15 cm B 5 cm 1 cm 15 cm.76 B 1.52 C 75 kn/m 7 75 kn/m D 25 kn/m kn/m E C 5 cm 1 cm 15 cm D 5 cm 1 cm 15 cm 3.81 F 75 kn/m kn/m G kn/m 1 25 kn/m E 5 cm 1 cm 15 cm F 5 cm 1 cm 15 cm G 75 kn/m 25 kn/m cm 1 cm 15 cm 1.67 Figure 2-7 Post-blast p-y curves for the east center pile of the 3x3 pile group at various depths during the first and tenth load series, where average R u is shown as a percent (Gerber, 23). 19

53 Soil Resistance (kn/m) Curve Depth Avg. Ru A. m 94% D 2.29 m 98% E 3.5 m 1% F 3.81 m 1% G 4.57 m 98% (B & C omitted, Avg. Ru < 9%) G F E D A Deflection (cm) Figure 2-8 Post-blast p-y curves for the east center pile of the 3x3 pile group during the first load series (Gerber, 23). Soil Resistance (kn/m) G F E D C B Curve Depth Avg. Ru A. m 63% B.76 m 56% C 1.52 m 7% D 2.29 m 77% E 3.5 m 65% F 3.81 m 66% G 4.57 m 51% A Deflection (cm) Figure 2-9 Post-blast p-y curves for the east center pile of the 3x3 pile group during the tenth load series (Gerber, 23). 2

54 2.5 Estimating P-Multiplier Adjustments for Diameter As part of the TILT Project Weaver (21) developed p-y curves for drilled shafts. The subsurface and loading conditions for the shafts and piles analyzed by Weaver and Gerber (2), respectively, are very similar. Because of these similarities coupled with the apparent lack of group effects in the pile group when the soil was fully liquefied, it is reasonable to assume the main difference between the p-y curves of Weaver and Gerber was the diameter of the foundation. Even though Weaver used different analysis procedures to derive his p-y curves when six p-y curves from the 324 mm steel pipe pile, were scaled using a 5.56 multiplier a good match was made with the.9 m drilled shaft. This suggests that a multiplier to account for differing foundation diameter can be applied to a general p-y curve equation for fully liquefied soils. Such an equation was provided by Rollins et al., (25) as being: p = A( By) C ( 2-1) where A = 3 x 1-7 (z + 1) 6.5, B = 2.8(z + 1).11, C = 2.85(z + 1) -.41, p is the soil pressure per length of pile (kn/m), y is the horizontal deflection (mm), and z is the depth (m). When the p-y curves for the.6 m diameter drilled shaft (Weaver et al., 25) were calculated a p-multiplier was also necessary. A multiplier of about 3.5 gave a good comparison. The error in this test could be greater due to the lower R u values. Based on the three different-diameter foundations in the TILT program, a equation to estimate the p-multiplier needed for diameter adjustments is p d = 3.81lnd (2-2) 21

55 where p d is the p-multiplier for diameter, and d is the diameter of the pile or shaft (m) Rollins et al., (25). 2.6 Static Test on Drilled Shaft MP-1 at the Mt. Pleasant Site The success of the TILT project has helped acquire additional funds to run several full-scale liquefaction load tests at a site near Charleston, South Carolina known as the Mt. Pleasant site. The soil profile has a liquefiable layer of silty sand that extends from 3.5 to 12 m below the ground surface. The sand is underlain by a stiff clay known as Cooper Marl. Hales (23) analyzed the static load tests performed on the drilled shaft designated MP-1 at the Mt. Pleasant test site. MP-1 was constructed to the same dimensions as was MP-3 (Figure 4-43). Three load tests were performed; the first was performed to evaluate the stiffness of the soil-pile system before liquefaction. The next two were performed after the soil was liquefied using downhole charges. The first load test in liquefied soil consisted of 1 different load cycles. After the fifth load cycle another set of charges was detonated. During the test the pore pressures would dissipate due to the time it took to load the test shaft, so the detonations were necessary to maintain a liquefied state. The second load test in liquefied soil consisted of a series of 7 load cycles. As with the first load test in liquefied soil, an intermediate blast was necessary to maintain a liquefied state. The intermediate blast was detonated after the fourth load cycle. Figure 2-11 and Figure 2-12 illustrate the peak pore pressures immediately after the blasts. 22

56 MP-1 test shaft was equipped with inclinometers, strain gauges along the length of the test shaft, and LVDTs at the pile head. Since the tests were static accelerometers were not installed. Figure 2-1 illustrates the load deflection curves produced by the three load tests performed on MP-1. Hales (23) computed p-y curves from strain gauge data along the length of the shaft. These p-y curves were compared to those produced during the analysis of the data from the TILT project mentioned in Sections 2.3 and 2.4. Hales found that a p-multiplier of 8 to adjust for diameter effects produced a reasonable fit with the p-y curves produced from the TILT project. Load (kn) Pre-blast -1 First Blast -2 Second Blast Deflection (mm) Figure 2-1 Applied load versus pile head displacement curves for all three load tests from testing on MP-1 (Hales, 23). 23

57 Ru % 2% 4% 6% 8% 1% 12% (a) Time = 5 seconds 2 Depth (m) Inner Ring Middle Ring Outer Ring 1 12 Figure 2-11 Peak pore pressure ratio with depth immediately after the first blast for MP-1, at the Mt. Pleasant test site. Ru % 2% 4% 6% 8% 1% 12% (a) Time = 5 seconds 2 Depth (m) Inner Ring Middle Ring Outer Ring 1 12 Figure 2-12 Peak pore pressure ratio with depth immediately after the second blast for MP-1, at the Mt. Pleasant test site. 24

58 2.7 Brief History of the Statnamic Device (Bermingham 2) The use of the statnamic device got its start in Hamilton in The first proposal for the device was written in In 1987, through a joint effort by Berminghammer and TNO, the idea and concept of the statnamic loading system was refined. One main reason for the development of the statnamic load testing was the limited use of the dynamic test. If the large weight used to load the pile failed to fully mobilize the pile, a larger one was needed. The problem with that is that there is a point where the pile receives too much damage to make the test worth while. Additionally, after examining the data received from bolt-on strain gauges used on test shafts, a problem with their accuracy was found. Due to these two difficulties with the dynamic load test, coupled with the demand for a faster and less expensive way of testing in-situ piles, the need for a device like the statnamic was sparked. Statnamic loading systems were originally referred to as inertial load testing. A desired load is chosen and the reaction of the pile is then monitored by the instrumentation (i.e. strain gauges, accelerometers, and LVDTs). These load systems were first developed to be able to fully mobilize a capacity of 6 tons. By April of 1988, the first testing model was built in Hamilton, Ontario, Canada to evaluate the feasibility of accelerating a mass off the top of a pile. In May of 1988 the first tests with the model were successfully performed. At the time, the loading direction of the statnamic device was in a vertical plane. A second model was made and sent to TNO in Holland to develop the testing instrumentation specifically for the statnamic testing. With time, the development and use of the statnamic device grew in popularity, as did the need for larger loads. Currently, the largest statnamic device weighs in at 6MN. 25

59 In 1995 a hydraulic catching system was developed that permits 1 different piles to be tested with multiple load-cycles in each test. On a Federal Highways Administration project the first lateral statnamic test was performed in 1994 in Newbern, North Carolina. Up until this point the statnamic device had been used primarily for axial loads. In 1998 the statnamic system was used to apply an 8 ton lateral load over water on a 6-pile group in Mississippi. 2.8 Statnamic Test on Drilled Shaft MP-3 at the Mt. Pleasant Site. Dan Brown (2), was responsible for the statnamic load test report for the Mt. Pleasant site. In his analysis of MP-3, he used a single degree of freedom model to represent the soil-pile system. The inertial force was calculated assuming that the pile would act like a cylinder rotating about its base. By taking the mass moment of inertia and multiplying it by the rotational acceleration in relation to the displacement, the force due to inertia was calculated. The damping force was calculated by expressing the damping constant as a damping ratio. An assumed mass was used with a logarithmically decaying stiffness. As the stiffness was decreased to a constant value, the damping ratio also decreased to a constant value (Brown, 2). The result was a linear static equivalent soil resistance. The damping ratios calculated for the three tests were 35%, 46%, and 46% respectively for the first, second, and third load test. A comparison of these results with those from the analysis in this thesis will be made latter in Section

60 2.9 Current Research Focus The focus of this thesis is the analysis of the three tests performed on the 2.59 m diameter drilled shaft designated MP-3 subject to statnamic lateral loading in blast induced liquefied soil. The results from the statnamic tests will first be used to compare and evaluate the results from the aforementioned conclusions from Gerber s work on the TILT project. Since the diameter of the drilled shaft used in the statnamic tests was about eight times greater than those in the TILT project, the effects of stiffness and pile diameter on the measured pile response can be evaluated. Then the analysis will be compared to existing methods for calculating p-y curves. 27

61

62 3 Site and Soil Description 3.1 Site Location and Bridge Description The Mt. Pleasant site, where construction and testing of foundation MP-3 occurred, is located near the banks of the Cooper River where a new bridge was constructed. Figure 3-1 is an aerial photograph that shows the location of the test site and the location of the new bridge. Figure 3-2 provides a drawing showing a closer view of the test site. This site, in addition to two others, was set aside for the construction and testing of prototype drilled shaft to provide site-specific data needed in the design of the new Cooper River Bridge, recently named the Arthur Ravenel, Jr. Bridge. The new bridge was dedicated and opened to traffic on July 16, 25. The Ravenel Bridge has a clear span of 471 m (1546 feet), making it the longest cable-stayed span in North and South America. This modern-looking bridge has also been designed with a vertical clearance of 56.7 m (186 feet) and a deck width of 39.3 m (129 feet). The two towers are each m (55 feet) tall, and the bridge will run a total length of 3.7 km (2.5 miles). Figure 3-3 is an artist s rendering of the new cable-supported Ravenel Bridge that is replacing the two existing truss bridges. 29

63 Figure 3-1 Aerial photograph of Cooper River bridges and test site. (taken from a presentation by the SCDOT and S&ME to the CE Club). 3

64 (a) (b) ' 15 3 Approximate location of test site LTB-1 Figure 3-2 Site map showing approximate locations of SPT and CPT borings (a) relative to the existing bridge approach ramps and (b) relative to the test site (Brown, 2). 31

65 Figure 3-3 Artist's rendering of future Ravenel Bridge (Bridgepros, 25). 3.2 Geological Background The soils at the test site are generally composed of alluvial silty sands and sands from the ground surface to a depth of about 12.5 m underlain by the Cooper Marl. Groundwater is generally present at a depth ranging from near the ground surface to a depth of 1.5 m, depending on tidal fluctuations. The sandy sediments of the coastal plain are typically loose, uncemented Pleistocene-age materials which are reported to have liquefied in the Charleston earthquake of (Elton and Hadj-Hamou, 199). The Cooper Marl is an Eocene to Oligocene-age marine deposit, described as a fossiliferous micrite or a soft, very fine-grained impure carbonate deposit (Heron, 1968; Malde, 1959). The formation typically consists of 25 to 75% carbonates, 1 to 45% very fine sand, 2 to 32

66 5% clay, and 5 to 2% phosphate (Heron, 1968). The calcium carbonate particles are typically very fine (<.2 mm) according to Heron (1968) and Malde (1959). 3.3 Scope of Geotechnical Investigation Prior to designing the bridge, a comprehensive geotechnical investigation was carried out to define the characteristics of the subsurface materials at the site. Preliminary investigations were initially performed by Parson-Brinkerhoff and more detailed investigations were performed by S&ME, Inc. The geotechnical investigation consisted of conventional sampling and laboratory testing as well as in-situ testing. Conventional sampling included undisturbed samples obtained with a thin-walled Shelby tube sampler, as well as disturbed soil samples obtained with a standard (5 mm OD) split-spoon sampler. Laboratory testing was performed on many field samples to determine particle size distribution, Atterberg limits, soil classification, shear strength and consolidation characteristics. In-situ tests included standard penetration (SPT) testing, cone penetrometer (CPT) testing, and shear wave velocity testing. The locations of the various test holes relative to the test pile groups are shown in Figure Test Borings and Laboratory Investigations Three test holes were drilled and sampled near the test site, namely DS-1, MPS- 11, and LB-28. Partial test-hole logs for these three borings are presented in Figure 3-4 through Figure 3-6. The test holes were advanced using rotary mud drilling. 33

67 Undisturbed samples of the Marl were obtained by pushing a 76.2 mm diameter, thinwalled Shelby tube using the hydraulic rams on the drill rig. Disturbed samples of the cohesionless soil were obtained using a standard 5.8 mm diameter split-spoon sampler with both donut and safety hammers. The type of hammer and sampler used is indicated on the test-hole logs. Each sample obtained in the field was classified in the laboratory according to the Unified Soil Classification System (USCS). Mechanical (sieve) analyses were performed on a number of the disturbed samples; and for these cases, the percentage of fines (material less than #2 sieve) is shown on the boring log. Atterberg limits, (plastic limit [PL], liquid limit [LL], and plasticity limit, [PI]) and natural moisture contents were determined for many undisturbed samples of the Cooper Marl. The results are also shown on the boring logs. In addition, the fines content was obtained from hydrometer analyses of the Marl. Because the behavior of the Marl was relatively unknown, both undrained and drained shear strength parameters were determined. Undrained strength was obtained from UU and CU triaxial shear tests while drained strength parameters were obtained from CU triaxial tests with pore pressure measurements. Based on the test hole logs, Camp et al., (2a) developed an idealized soil profile for the site. This profile, consisting of six layers with some minor modifications, is shown in Figure 3-7. The first layer typically extends from the ground surface to a depth of 1.5 m and consists of loose, poorly graded fine sand (SP) to silty sand (SM). In some cases, sandy clay layers were interbedded in this material. The surface sand was typically underlain by a sandy clay layer 1. to 1.5 m thick, which classified as CH material. This clay layer was very soft and had an average natural moisture content of 34

68 about 16%, which is approximately the same as the liquid limit, suggesting that the clay is normally consolidated. The PI was typically about 7%. The third layer was also a loose, fine sand (SP) to silty sand (SM) similar to the first layer and typically extended to a depth of 5.5 m. The fines content varied considerably with depth and from hole to hole with a range from.5 to 28%. The fourth layer was typically located between 5.5 and 8.5 m below the ground surface. This layer was also a sand but contained significantly more fines. The layer typically classified as silty sand (SM) or clayey sand (SC). The natural moisture content was 3% and the fines content varied from 15 to 24%. The fifth layer generally began at 8.8 m depth and extended to the top of the Cooper Marl. This layer contained fewer fines and was generally classified as a loose to medium dense poorly graded fine sand (SP). The Cooper Marl was first encountered between 12.5 and 14 m below the ground surface and extended to a depth of 85 m, which was below the base of all the test foundations at the site. The Cooper Marl is a stiff, high plasticity calcareous silt or clay. The results of Atterberg limit testing are plotted on a plasticity chart in Figure 3-8, and the Marl generally classifies as MH or CH material (Camp et al., 2a). The liquid limit typically ranges from 5 to 9% with a plasticity index varying from 15 to 6%. However, in some cases, the liquid limit was approximately 135% with PIs varying from 1 to 8%. The moisture content and fines content in the Cooper Marl are plotted in Figure 3-9 and Figure 3-1 (Camp et al., 2b). The natural moisture content is typically between 4 and 6%, which is somewhat higher than the plastic limit but much lower than the liquid limit, suggesting that the Cooper Marl is overconsolidated. 35

69 The undrained shear strength is plotted as a function of elevation in Figure 3-11 (Camp et al., 2b). The Marl is very stiff with undrained strengths typically ranging from 1 to 2 kpa at the top of the layer and increasing with depth to a value between 2 and 3 kpa at a depth of about 45 m. Below this depth, the strength appears to remain relatively constant. The results of the drained shear strength tests on the Cooper Marl are plotted in the form of a p-q diagram in Figure 3-12 (Camp et al., 2b). The best-fit line through the data points indicates that the drained friction angle is approximately 43 with relatively little variation about the line. 36

70 Figure 3-4 Boring log for test hole DS-1 (Hales 23). 37

71 Figure 3-4 (Continued) Boring log for test hole DS-1. 38

72 Figure 3-4 (Continued) Boring log for test hole DS-1. 39

73 Figure 3-4 (Continued) Boring log for test hole DS-1. 4

74 Figure 3-4 (Continued) Boring log for test hole DS-1. 41

75 Figure 3-5 Boring log for test hole MPS-11 (Hales 23). 42

76 Figure 3-5 (Continued) Boring log for test hole MPS

77 Figure 3-6 Boring log for test hole LB-28 (Hales 23). 44

78 Figure 3-6 (Continued) Boring log for test hole LB

79 Figure 3-6 (Continued) Boring log for test hole LB

80 Figure 3-6 (Continued) Boring log for test hole LB

81 Figure 3-6 (Continued) Boring log for test hole LB

82 Sand (SP) to Silty Sand (SM), Loose, fine, Avg. N=5 Sandy Clay (CH), soft, w=16, LL=14, PI-69 5 Sand (SP), loose, fine, Avg. N=6,.5 to 28% Fines Depth (m) 1 Silty Sand (SM) and Clayey Sand (SC) Avg. N=7, w=3% Sand (SP), fine, loose to medium dense, Avg. N=12 15 Cooper Marl (CH) stiff to very stiff Avg. N=15, 4%<w<5% 5%<LL<15%, 2%<PI<8% 2 Figure 3-7 Idealized soil profile for the Mt. Pleasant test site (Modified from Camp et al., 2a). 49

83 Figure 3-8 Atterberg limits tests at various depths within the Cooper River Marl relative to the plasticity chart (Camp et al., 2b). Figure 3-9 Natural moisture content versus elevation in the Cooper River Marl (Camp et al., 2b). 5

84 Figure 3-1 Fines content versus elevation in the Cooper River Marl (Camp et al., 2b). Figure 3-11 Undrained shear strength versus elevation from UU and CU triaxial shear test on undisturbed samples of the Cooper River Marl (Camp et al., 2b). 51

85 Figure 3-12 Results of drained triaxial shear strength tests on Cooper River Marl plotted in a p-q diagram (Camp et al., 2b). 3.5 In-Situ Testing Standard Penetration (SPT) Testing Standard penetration (SPT) testing was performed at many locations within the sandy alluvial deposits above the Cooper Marl. Testing was performed by dropping a N weight from a height of.762 m. The raw standard penetration value (N) is the number of blows required to drive the split-spoon sampler through.3 m of penetration. For test boring LB-28, the sampler was driven.6 m and the N value was taken from the middle.3 of the interval. The (N 1 ) 6 value was determined using the equation ( N 1 ) 6 = NC C n E = q c Pa ' σ vo.5 E 6% applied (3-1) 52

86 where C E is the correction for the percent energy applied and E applied is the percentage of the theoretical energy applied by the hammer, P a is atmospheric pressure, and σ vo is the vertical effective stress. Finally, the equivalent clean sand blow count [(N 1 ) 6-CS ] was obtained using the equation ( N1 ) 6 CS = α + β ( N1) 6 (3-2) where α = exp(1.6 (19/FC 2 ), β = [.99 + (FC 1.5 /1)], and FC = Fines Content (%) for fines contents between 5 and 35%. For FC less than 5%, α = and β =1; and for FC greater than 35%, α = 5 and β = 1. The (N 1 ) 6-CS for each test in each test hole is plotted as a function of depth in Figure The relative density was computed using the equation.5 ( N1) 6 CS D r = (3-3) 4 presented by Kulhawy and Mayne (199). The relative density is plotted as a function of depth for each of the three test holes in Figure The average relative density (5%) and standard deviation bounds (± 14%) are also shown in Figure Although there is considerable variation in relative density within the various test holes, the average and range of values stay relatively constant with depth. 53

87 (N 1 ) 6cs Depth (m) LB-28 DS-1 MPS Figure 3-13 Normalized SPT clean sand penetration resistance versus depth for three test holes near the test site. 54

88 Relative Density, Dr (%) LB-28 DS-1 MPS-11 Depth (m) Mean - One Std. Dev. Mean Mean + One Std. Dev. 18 Figure 3-14 Interpreted relative density versus depth based on SPT penetration resistance for three holes close to the test site. 55

89 3.5.2 Cone Penetration (CPT) Testing Three cone penetration (CPT) soundings (MPS-7, LTB-1 and GT-1) were performed at several locations near the test area, as shown in Figure 3-2. Two of the soundings were performed by S&ME, while the third (GT-1) was performed by Georgia Tech researchers. The CPT performed by S&ME used a track-mounted cone rig equipped with an automated data acquisition system. The cone was a piezocone with a 1 cm 2 surface area. The Georgia Tech cone rig used a small truck-mounted system with a seismic piezocone with a 15 cm 2 surface area. The porous filter for the both cones was located in position 2, approximately 12 mm from the tip. The tests were conducted in general accordance with ASTM D The soundings varied in depth from 13 to 3 m. The normalized cone (tip) resistance (q c1 ), friction ratio (f r ) and pore water pressure (u) for each of the soundings are presented as a function of depth below the ground surface in Figure 3-15 through Figure The CPT results were used to interpret the soil profile using the correlation with soil behavior type developed by Robertson and Campanella (1988). The soil profile interpreted from the CPT soundings is also shown for each sounding. Although the soil profile is generally similar for each sounding and matches the idealized profile identified from the three test borings, there is still considerable variation in the magnitude of tip resistance for the three soundings. The presence of the clay layers in the profile is clearly indicated by a decrease in the tip resistance, an increase in the friction ratio and an increase in the dynamic pore pressure above the static pressure line (U o ). In addition to the clay layer between 1.5 and 3.5 m and the Cooper Marl layer, the sand layer generally between 5.5 and 1 m appears to 56

90 contain a relatively high clay content, based on an increase in friction ratio and pore pressure Interpreted Soil Type Silty Sand (SP to SM) Sand (SP) Clay (CH) Clay (CH) Sand (SP) Clayey Sand (SC) Sand (SP) Cooper Marl (CH) Tip Resistance qc1 (MPa) Friction Ratio f r (%) Pore Pressure U2 (MPa) U2 U Relative Density Dr (%) Depth (m) Figure 3-15 Results from CPT sounding LTB-1 including normalized cone resistance, friction ratio, and pore pressure along with interpreted relative density and soil profile. 57

91 Interpreted Soil Type Sand (SP) to Silty Sand (SM) Clay (CH) Sand (SP) Silty Sand (SM) and Clayey Sand (SC) Sand (SP) Cooper Marl (CH) Tip Resistance qc1 (MPa) Friction Ratio fr (%) Pore Pressure U (MPa) U2 U Relative Density Dr (%) Depth (m) Figure 3-16 Results from CPT sounding MPS-7 including normalized cone resistance, friction ratio, and pore pressure along with interpreted relative density and soil profile. 58

92 Interpreted Soil Type Sand (SP) to Silty Sand (SM) Clay (CH) Sand (SP) Silty Sand (SM) and Clayey Sand (SC) Sand (SP) Cooper Marl (CH) Tip Resistance qc1 (MPa) Friction Ratio fr (%) Pore Pressure U (kpa) U2 U Relative Density Dr (%) Depth (m) Figure 3-17 Results from CPT sounding GT-1 including normalized cone resistance friction ratio, and pore pressure along with interpreted relative density and soil profile. 59

93 Relative Density Based on CPT The relative density (D r ) of the coarse-grained layers was estimated from the CPT cone resistance using the equation.5 q c1 pa D r = (3-4) 35 developed by Kulhawy and Mayne (199), where p a is atmospheric pressure, q c1 is the cone resistance at a vertical effective stress of one atmosphere, and the sand is assumed to be normally consolidated. The q c1 value is given by the equation p a qc 1 = qccq = qc ' σ vo.5 (3-5) where σ vo is the effective vertical stress and the adjustment factor C Q is less than or equal to 1.7. The friction angle in the granular layers was estimated using an equation developed by Bolton (1986). The triaxial compression friction angle (φ tc ) is given by the equation φ = φ + 3I (3-6) tc cv rd where φ cv is the critical void ratio friction angle. Bolton (1986) recommends a value of 31 to 33 degrees for φ cv in quartz sands with some silt and we have assumed a value of 31 in this study. I rd is given by the equation 6

94 p f I 1 ln 1 rd = Dr 1 (3-7) pa where p f is the mean effective stress at failure. Horizontal pressures at the failure state were estimated using Rankine values for active and passive pressures. The relative density and friction angle versus depth profiles computed using Equation 3.4 and Equation 3.6, respectively, are shown for each test hole in Figure There is a substantial variation in relative density and friction angle in the surface sand layer; however, in the sand layer immediately below the surface clay, there appears to be reasonable agreement. In this fine sand layer, the relative density is approximately 5% with a friction angle of 38. In the underlying silty to clayey sand, the relative density and friction angle drop significantly relative to the cleaner sand above, and the variability also increases. In this clayey sand layer, the relative density appears to vary from 2 to 5% with an average of about 4%, while the friction angle generally varies from 31 to 35 with an average of about 33. In the lowest fine sand layer, the interpreted values are once again more consistent. The average relative density is about 5% with a friction angle of 3l.5. 61

95 Interpreted Soil Type Sand (SP) to Silty Sand (SM) Relative Density D r (%) Friction Angle φ ( degrees) Clay (CH) MPS-7 GT-1 LTB-1 MPS-7 GT-1 LTB-1 5 Sand (SP) 5 5 Depth (m) Silty Sand (SM) and Clayey Sand (SC) Sand (SP) 15 Cooper Marl (CH) Figure 3-18 Interpreted Relative density and friction angle versus depth for sand layers in the soil profile based on three CPT soundings. Undrained Shear Strength Interpretation Based on CPT The undrained shear strength of the fine-grained layers in the profile was estimated from the CPT cone tip resistance using the equation S u q = c σ o N k (3-8) where q c is cone tip resistance, σ o is the total vertical stress, and N k is the bearing capacity factor for an electric piezocone. According to Robertson and Campanella (1988), the N k value typically ranges from 1 to 2 and was assumed to be equal to 15 for this study. Although the undrained shear strength obtained from Figure 3-19 is only an estimate, the 62

96 Interpreted Soil Type Undrained Strength Su (kpa) Sand (SP) to Silty Sand (SM) Clay (CH) 5 Sand (SP) 5 1 Silty Sand (SM) and Clayey Sand (SC) 1 MPS-7 GT-1 LTB-1 Sand (SP) Depth (m) 15 2 Cooper Marl (CH) Figure 3-19 Interpreted Undrained shear strength versus depth for clay layers in the soil profile based on three CPT soundings. approach does provide a continuous profile that shows the consistency of the strength within layers in the profile. The undrained shear strength computed using Equation 3-8 is shown as a function of depth in Figure The undrained strength in the upper clay layer ranges from 25 to 5 kpa and tends to increase with depth. The estimated 63

97 undrained strength in the Cooper Marl also tends to increase with depth with an average strength of about 125 kpa at the top of the layer and about 275 kpa at the bottom. This trend and the strength values shown in Figure 3-19 are generally in good agreement with the measured undrained shear strength from laboratory tests shown in Figure Shear Wave Testing The shear wave velocity (V s ) profile was measured during two seismic cone penetrometer (SCPT) soundings (MPS-7 and GT-1), which were described previously, and one conventional down-hole test performed in test hole DS-1 by Bruce Redpath of Redpath Geophysics. The shear wave velocity was normalized for overburden pressure (V s1 ) using the equation V s 1 = V s 1 ' σ o.25 (3-9) The shear wave velocity profiles obtained from the two SCPT tests and the downhole test are provided in Figure 3-2. Figure 3-2 (a) presents the V s results, while (b) presents V s1 to a depth of 3 m. The agreement between the three tests is relatively good. In some cases where discrepancies occur, the source appears to be an over-estimate of velocity at one depth followed by an underestimate at the subsequent depth, so that the average is about the same as in the other tests. In many cases, the V s1 value is greater than 2 m/sec, indicating that the sand is not susceptible to liquefaction; however, the SPT and CPT values indicate that liquefaction is likely to occur, as discussed subsequently. The V s value in the Marl is higher than in the overlying sands and increases progressively from 335 m/sec at the top of the layer to 762 m/sec at the base. However, when the correction is applied to get V s1, the velocity in the Marl remains 64

98 nearly constant, suggesting that the material properties remain essentially constant with depth. Shear Wave Velocity V s (m/s) Shear Wave Velocity V s1 (m/s) GT 5 LTB-1 Redpath 5 GT LTB-1 Redpath Figure 3-2 Profiles of V s and V s1 versus depth based on down SCPT sounding and a down-hole shear wave velocity test conducted by Redpath Geophysics. 3.6 Liquefaction Hazard Analysis On August 31, 1886 a major earthquake struck Charleston, South Carolina. Because no instrumental records are available, it is not possible define the magnitude and peak acceleration precisely, however, Bollinger (1977) has assigned a magnitude of

99 and an intensity of X on the modified Mercalli scale to the event. The earthquake caused an estimated $5 to $6 million damage to buildings, (see damage in Figure 3-21), as well as liquefaction in the subsurface materials (Stover and Coffman, 1993). A photo of a large sand boil produced by liquefaction during the 1886 earthquake is shown in Figure For an excellent compilation of first-hand observations of the 1886 Charleston earthquake, the reader is referred to Peters and Herrmann (1986). The new Ravenel Bridge has been designed to withstand an earthquake magnitude similar to the earthquake that struck the area in 1886, as well as smaller events which might occur more frequently. Despite research by a number of investigators, the exact fault mechanism responsible for the 1886 Charleston earthquake is still not known. Because there was no surface manifestation of the fault, geophysical techniques have been employed to identify potential faults; however, no consensus has developed within the scientific community regarding the causative fault. Instead, a number of areal source zones have been developed to quantify the earthquake hazard (Elton and Hadj-Hamou, 199). Based on these source zones, the U.S. Geological Survey has developed probabilistic ground motion estimates for the Charleston area (Frankel et al., 2). Maps are available for acceleration levels corresponding to 2%, 5%, and 1% probabilities of exceedance in 5 years. These acceleration levels correspond to return periods of approximately 25, 1 and 5 years. For the design of bridges in South Carolina, the South Carolina Department of Transportation requires a two level design approach. For more frequent small to moderate earthquakes, bridges are designed so that the bridge will remain essentially elastic and can remain in full service with little if any damage. For this first-level event, known as the Functional Evaluation Earthquake 66

100 (FEE), the peak ground acceleration values are associated with a 1% probability of exceedance in 5 years. The bridge is also designed for a major earthquake, known as the Safety Evaluation Earthquake (SEE). This earthquake is used to ensure that the structure does not collapse and that there is no loss of life. In addition, for critical bridges, service should be maintained and damage should be easily detectable and repairable. The acceleration level for the SEE event is associated with a 2% probability of exceedance in 5 years. According to the US Geological Survey, the peak ground acceleration at the bridge site with a 2% probability of exceedance in 5 years is.77g and is associated with a modal magnitude 7.3 earthquake. The peak ground acceleration at the bridge site with a 1% probability of exceedance in 5 years is.16g and is associated with a modal magnitude 6.4 earthquake. A liquefaction analysis has been performed for the two earthquake levels described previously. The analyses were performed in accordance the recent recommendations of the MCEER workshop (Youd et al., 21). The analyses were performed using the CPT sounding identified as GT-1. The results of the analysis for the M7.3 and.77g event are shown in Figure For this relatively large event, the analyses indicate that essentially all of the profile above the Cooper Marl would liquefy, with the exception of the clay layer in the upper part of the soil profile. Despite the high fines content in the clayey sand layer, liquefaction would also be expected in this layer. The results of the analysis for the M6.4,.16g event are shown in Figure For this relatively small event, the analyses indicate that most of the sand above the Cooper Marl would not liquefy. Additional analyses, not shown here, indicate that this 67

101 event represents the upper bound for which liquefaction does not occur. If the acceleration or earthquake magnitude is any higher, then the entire sand layer above the Cooper Marl liquefies. In contrast, liquefaction analyses based on the shear wave velocity indicate that liquefaction would unlikely occur in the zone from 5 to 13 m below the ground surface because the shear wave velocity is greater than 215 m/sec. Figure 3-21 Photograph of a brick house wrecked by the Charleston earthquake of August 31, 1886 (USGS, 25). 68

102 Figure 3-22 Photograph of a sand boil due to liquefaction during the 1886 Charleston, South Carolina Earthquake (FHWA, 25). 69

103 Tip Resistance (MPa) SBT Index F.S. Against Liquefaction Relative Depth (m) Figure 3-23 Profiles showing cone tip resistance, SBT index, and factor of safety against liquefaction versus depth for GT-1 due to M7.3 earthquake producing.77 g peak acceleration associated with a 2% probability of exceedance in 5 years. (Hales, 23) 7

104 Tip Resistance (MPa) SBT Index F.S. Against Liquefaction Relative Depth (m) Figure 3-24 Profiles showing cone tip resistance, SBT index, and factor of safety against liquefaction versus depth for GT-1 due to M6.4 earthquake producing.16 g peak acceleration associated with a 1% probability of exceedance in 5 years. (Hales 23) 71

105

106 4 Test Set-Up and Pile Description 4.1 Introduction Three successively larger lateral statnamic load tests were performed on the pile designated MP-3 on August 29, 2, each about one hour apart. This testing was performed as part of the design phase for the replacement bridge for the Cooper River Bridge. Embedded or down-hole charges were used to induce high excess pore water pressure, causing liquefaction of the soil directly surrounding the test site. The following chapter provides an overview of the foundation construction, foundation properties, statnamic load test configuration, and the test layout. 4.2 Pile Description According to the testing report (Brown, 2), the construction of the pile numbered MP-3 was finished on July 29, 2. The consulting engineer on the project was S & ME, Inc. The test foundation was a reinforced concrete shaft with a permanent steel liner, also known as cast-in-steel-shell (CISS) pile. The outside diameter of the steel shell was 2.59 m with a wall thickness of 25.4 mm and extended to a depth of about 2.12 m below the ground surface. After the steel casing was installed using a vibratory hammer, the inside was drilled out and advanced to a final depth of 31.7 m prior to 73

107 pouring the reinforced concrete core. Figure 4-1 shows the drilling process used to construct MP-3. Figure 4-1 Contractor used a track-mounted SoilMec for drilling (photograph from a presentation by the SCDOT and S&ME to the CE Club). The average 3-day compressive strength of the concrete placed in the steel shell of MP-3 was 42.7 MPa (6.2 ksi) based on six laboratory-cured samples. The vertical reinforcement consisted of 36 #18 bars evenly spaced around a ring with a diameter of 2.14 m. Number 6 bar spiral reinforcement was used as the containment reinforcement with a pitch of 89 mm. A 15 mm cover was maintained between the spiral reinforcement and the inside of the steel casing. Figure 4-2 provides a photograph of the reinforcement cage construction for MP-3. 74

108 Figure 4-2 Photograph of worker assembling reinforcement cage at the Mount Pleasant site (photograph from a presentation by the SCDOT and S&ME to the CE Club). The steel casing was installed with a stick-up height of 2.13 m above the ground surface and extended to 2.12 m below the ground surface. After the installation of the steel shell, the verticality of the shaft was assessed by Trevi Icos Corporation. The test shaft was actually tilted about 1.1 degrees from vertical to the North-East. A drawing of the measured shaft alignment is presented in the Appendix. The concrete portion of the test shaft was then installed with a stick-up height of 1.83 m above the ground surface and extended to a depth of 31.7 m below ground. Figure 4-43 presents the pile dimensions. Figure 4-5 shows the drilled shaft together with the idealized soil profile. 75

109 Figure 4-3 Drilled shaft dimensions, strain gauges, and accelerometers. Figure 4-4 Drilled shaft dimensions, strain gauges, and accelerometers. 76

110 Figure 4-5 Drilled shaft and corresponding soil profile. 4.3 Test Set-Up The lateral load was applied with a statnamic device on a sled. Figure 4-6 provides a schematic of the test set up. The sled has a series of donut shaped weights that encase a core called the silencer cylinder. The vertical lines on the statnamic device schematic are the separate weights that give the device its mass and in turn its force. The statnamic sled, which weighs approximately 9.7 X 1 4 kg, was supported by a steel trestle with a timber crane mat deck. The trestle was supported by driven H-piles. The load was applied about 1.31 m above the ground surface through a hemispherical bearing 77

111 to allow rotation of the shaft, ensuring a free-head condition. A load cell at the point of load application was used to directly measure applied force. Figure 4-6 Schematic of statnamic load test at the Mt. Pleasant site (drawing modified from the Ravenel Bridge Project Load Test Plans). 78

112 4.4 Above Ground Instrumentation Two.76 m linear variable differential transducers (LVDT) were installed to directly measure the pile head deflection, one above the load point and the other below it. The LVDTs were mounted on an independent reference beam supported by driven piles within isolation casings. The top LVDT was mounted 1.65 m above the ground surface; the lower LVDT was located 1.1 m above the ground surface. Two accelerometers were mounted at the height of the load point on the opposite side of the load application. One of these accelerometers failed to give reliable readings, so the data was not used for analysis. Two more accelerometers were mounted above ground. The first was mounted above the load point at a height of 1.89 m. The other was mounted on the reference beam to detect any significant movement of the structure that could affect the LVDT data. The location of the instrumentation is shown in Figure

113 Figure 4-7 Schematic of the statnamic loading device with accelerometers and LVDTs at the Mount Pleasant site (Figure provided by AFT Inc. load report for MP-3). 8

114 4.5 Below Ground Instrumentation A series of eight accelerometers or Down-hole Lateral Motion Sensors (DLMS) were installed with individual guide mounts that were lowered into a grooved inclinometer casing that was pre-cast into the test shaft. The accelerometers were installed to measure lateral motion in the direction of the load test. The accelerometers were placed at the following elevations (meters below the ground surface): 1.83, 3.83, 5.83, 7.83, 9.83, 11.52, 13.22, as shown in Figure Strain gauges were also installed a ten depth stations to measure the bending in the pile. Resistance-type strain gauges were mounted on a separate rebar consisting of a 4 foot long #4 bar tied to the rebar cage before installation. For the first 5 stations two gauges placed in line with the direction of loading with a horizontal distance between gauge pairs of 2.14 m. For stations 6-1, an additional two strain gauges were placed perpendicular to the direction of load application and similarly spaced. The depths and layout of the strain gauges are shown previously in Figure All of the data (load, deflection, acceleration, and strain) were collected using a Megadac data acquisition system at a sampling rate of 2 samples per second for each channel. A pre-trigger of.5 seconds before the loading was used to start recording the data while a 5 second post-trigger was used to terminate data acquisition. Piezometers were installed in the ground around the test shaft to quantify the pore water pressure increase induced by the explosive charges, along with subsequent dissipation of pore pressure. Plan and profile views of the piezometer layout relative to the test shaft and the explosive charges are presented in Figure 4-9 and Figure 4-1, respectively. The transducers were basically arranged to produce three vertical arrays of 81

115 sensors at radial distances of 1.83, 7.32, and 1.36 m from the center of the test shaft. The vertical arrays were typically spaced at about 1.5 m depth intervals. The piezometer number and depth below ground surface is indicated in this figure. Figure 4-8 Reinforcement cage after installation of strain gages and inclinometers (photograph from a presentation by the SCDOT and S&ME to the CE Club). The piezometers labeled with a B designation consisted of piezoresistive transducers designed to withstand transient blast pressures of up to 41 MPa (6 psi) while still resolving the residual pore pressure to within ±.7 MPa (.1 psi). The transducers with an A designation consisted of electrical resistance transducers with a resolution of 6.9 kpa (1 psi) when used in conjunction with the Megadac data collection system. The A-type transducers experienced an unusually high failure rate during field testing and typically only provided marginally useful data; however the B-type sensors generally performed well. 82

116 3.96 m R 4.57 m R 5.18 m R A5? m B15 A4 6.4m 1.83m B m A3 4.88m A2 1.67m B11 B1 B12 B m 3.35m B9 6.4m 4.88m A1 7.93m 1.6m B4 B5 B3 3.2m 6.4m B2 1.67m MP-3 B m 1.83m B6 4.88m B7 7.93m B8 1.98m Load Direction BYU Piezometer AFT Piezometer Blast Holes Inner Ring 1.83 m R Middle Ring 7.32 m R Outer Ring 1.36 m R Figure 4-9 Plan view of the piezometers and charges (Brown, 2). 83

117 Figure 4-1 Elevation view showing a profile of piezometers and down-hole charges relative to the test shaft. 4.6 Blast Layout Liquefaction is manifest by an increase in pore water pressure. This increase in pore water pressure was achieved by detonating explosive charges distributed around MP-3 test shaft. A pilot liquefaction test was performed at a location separate from the test site for MP-3. The pilot test was done to better define the charge weight, spacing, and delay sequence so as to fully liquefy the soil around the test shaft. For each of the 84

118 load tests, eight charges spaced evenly around three different radii were detonated. The first eight blast holes were spaced around a radius of 3.96 m. The second and third blast rings radii were 4.57 m and 5.18 m. The charges were staggered to help avoid sympathetic detonations. The first blast series set a.68 kg (1.5 lb) charge at 3.5, 6.1, 9.14, and m below the ground surface. The second blast series used.91 kg (2 lb) charges at 4.57, 7.62, and 1.67 m depths. The third and final blast series put.68 kg (1.5 lb) charges at the same depths as in the first blast series. This layout can be seen in Figure 4-9 and Figure 4-1. Even though the charges were staggered, some sympathetic blasts still occurred during the second blast series at 9.14 and m depths. The binary explosives consisted of a mixture of ammonium nitrate and nitromethane, and the weights are given in equivalent weights of TNT. During each of the three blast sequences, the charges were detonated two at a time with a delay of 25 msec between detonations. The charges were detonated beginning around the bottom ring and then moved upward around each subsequent ring to the top. 85

119

120 5 Statnamic Lateral Load Test Results 5.1 Introduction To determine the effects of liquefaction on the lateral resistance of deep foundations, dynamic lateral load tests were performed on August 29, 2. Three lateral load tests were performed with a statnamic load sled. With each test, the load was progressively increased. Before each test, a set of down-hole explosives was detonated to increase the pore water pressure in the soil to induce liquefaction. The planned order of the load tests went as follows: 1. Detonate the embedded charges to liquefy the soil. 2. Wait about 45 seconds for dissipation of gasses from the explosions. 3. Fire the statnamic device. After each detonation of the down-hole charges and loading, there was evidence of sand boils and water pressure venting at the ground surface around the test site. Some settling also occurred after each test. After the third test was completed, the maximum surface elevation change was more then a quarter of a meter (see Appendix A). During the three lateral load tests, the maximum deflection obtained was more than 98 mm, and the largest load used was more than 65 kn. Data was recorded for more then 8 seconds for all three tests. Analysis of the data yielded no significant deflections after about 1.5 seconds following the test firing; therefore the data was truncated at this point, and all the analysis 87

121 of the data was done in that amount of time. The results of the three tests are described in the following sections of this chapter. 5.2 Lateral Load Tests First Lateral Load Test The first lateral load test, as was mentioned in chapter 4, had two LVDTs that measured the pile head deflection, one above the load point and one below. Since the actual deflection at the load point wasn t measured, linear interpolation was used to determine the deflection at the load point. There is some potential for error because of this linear assumption, but since the distance between the two LVDTs is very small, the error should be negligible. The time histories of the upper and lower LVDT deflections are presented in Figure 5-1. As shown in Figure 5-2, a maximum load of 45 kn was achieved, and the maximum deflection attributed to that load was 59.6 mm for the first test. During the first lateral load test there seemed to be some high frequency vibrations or noise. These vibrations can be seen in the load-time history just after.5 seconds in Figure 5-2. This noise is thought to be created by the initial reaction of the donut weights around the silencer cylinder. No matter how much the weights are pushed together, inevitably some slack persists. When the load is applied, this slack is eliminated, which causes the weights to impact the system, initially inducing these high frequency vibrations. Puffs of dust from in-between the load weights were also noticed just after the load was applied, evidence of slack that needed to be removed. The load 88

122 system had a low frequency as can be seen in the LVDT deflection time histories, so the high frequency vibration could not be from vibration of the total weight of the device, and this phenomenon was not noticed in the second and third tests, providing additional evidence that the vibration came from the elimination of the space between the weights upon loading. 7 Deflection (mm) Deflection Lower LVDT (mm) Deflection Upper LVDT (mm) Time (sec) Figure 5-1 Pile head deflection time history for the first lateral load test on test pile MP Load (kn) Load (kn) Time (sec) Figure 5-2 Load time history for the first lateral load test on test pile MP-3. 89

123 Load (kn) Deflection (mm) Figure 5-3 Load versus deflection curve for load test 1 on test pile MP Second Lateral Load Test The second load test data was a little cleaner, since the statnamic device had already been fired. In the second and third load tests the high frequency vibration is not evident in either the load or acceleration time histories. The LVDT deflections were zeroed using the average deflection obtained from the first thirty points of the first load test s LVDT data to maintain the residual deflections from that first test that were present at the start of the second test. In contrast to the deflection, the load at the beginning of each load test should be zero. Therefore, the load data were zeroed using the average of the first thirty points of the load. The maximum load obtained in the second test was almost 55 kn and the maximum deflection caused by this load was 87.4 mm. The deflection and load time histories for the second test are plotted in Figure 5-4 and Figure 5-5, respectively, while the load-deflection curve is plotted in Figure

124 Deflection (mm) Deflection Lower LVDT (mm) Deflection Upper LVDT (mm) Time (sec) Figure 5-4 Pile head deflection time history for the second lateral load test on test pile MP Load (kn) Load (kn) Time (sec) Figure 5-5 Load time history for the second lateral load test on test pile MP-3. 91

125 6 5 4 Load (kn) Deflection (mm) Figure 5-6 Load versus Deflection for load test Third Lateral Load Test The third load test data was zeroed in the same manner as was the second load test. The maximum load that the statnamic device delivered was just over 65 kn, and the resulting peak deflection was 98. mm. The deflection and load time histories for the third test are plotted in Figure 5-7 and Figure 5-8, respectively, while the load-deflection curve is plotted in Figure

126 1 Deflection (mm) Deflection Lower LVDT (mm) Deflection Upper LVDT (mm) Time (sec) Figure 5-7 Pile head deflection time history for the third lateral load test on test pile MP Load (kn) Load (kn) Time (sec) Figure 5-8 Load time history for the third lateral load test on test pile MP-3. 93

127 7 6 5 Load (kn) Deflection (mm) Figure 5-9 Load versus Deflection curve for load test Pile Motion from Acceleration Data The eight DLMSs or the downhole accelerometers can be used to determine acceleration, velocity and deflections along the length of the pile. To derive the velocity and displacement from the acceleration, the acceleration needed to be zeroed to start with no motion, as was true with the test. A baseline correction was also necessary to adjust for the small drift in the accelerometer during the test. Although the drift in baseline acceleration is very small, drift in the accelerometer during the test can still lead to unrealistic displacements after double integration over the time history. Baseline correction has been used widely in the arena of earthquake engineering to adjust the accelerations obtained from their instruments. A program called Baseline (Gregor, 21) was used to baseline correct the accelerations. This program performed a least squares inversion between an input deflection time history and an n th degree polynomial. The 94

128 accelerations could then be integrated with respect to time to derive the velocities, and again to get the deflections. These deflections were then compared to the LVDT deflections to ensure that the baseline correction was reasonable. The accelerometer attached to the reference frame showed very small movements, so small that they were considered insignificant for all three load tests. All of the acceleration, velocity, and deflection graphs show that the peak velocity occurs where the acceleration is zero, and the velocity is zero when the deflection is at its maximum as expected by calculus. This correlation between acceleration, velocity, and deflection is a good indication the accelerations were properly integrated. On the acceleration graphs the maximum values always occur as negative. As the pile is loaded, it has to push the soil out of the way. In the unloading portion of the graphs the gap the pile produced allows it to travel more freely causing the accelerations to be greater. The peak accelerations for all three tests are greater than those seen during an earthquake, but this isn t a problem since the inertial force can easily be removed from the measured force. Typical accelerations seen during an earthquake might range from.3 g to 1.3 g. The peak velocities however, are values similar to those seen during earthquakes, considering an average velocity from an earthquake is about 1 m/s/g. Therefore, the acceleration to which this loading corresponds to can be calculated based on the measured velocity. Also since velocity is proportional to the damping force, the damping for this test is likely to be similar to what is expected in an earthquake. 95

129 5.3.1 First Load Test Pile Motion from Acceleration Data Figure 5-1 through Figure 5-13 show the measured acceleration, and derived velocity and deflection time histories from each accelerometer location along the length of the test pile for the first load test. Almost all the accelerometers worked during the test; however, one accelerometer malfunctioned at the load point. Since two accelerometers were placed at that location, an acceleration time history was still recorded. They can be seen in Figure 5-1b. Figure 5-14 through Figure 5-16 show acceleration, velocity, and deflection versus depth profiles at several selected time steps to provide an idea of the pile motion during each test. The deflections from the two LVDTs are plotted along with the acceleration-derived deflections in Figure In general, the computed deflections agree reasonably well with the LVDT deflections, however. The main source of error is most likely from baseline correction of the accelerations. 96

130 Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) (a) 1.89 m (6.2 ft) above ground Time (sec) Velocity Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) (b) 1.31 m (4.3 ft) above ground Time (sec) Velocity Figure 5-1 Acceleration, velocity, and deflection graphs from test 1 accelerometers. 97

131 Acceleration & Deflection Time (sec) Acceleration (m/s^2) Deflection (mm) Velocity (m/s) (a) 1.83 m (6. ft) below ground Velocity Acceleration & Deflection Time (sec) Acceleration (m/s^2) Deflection (mm) Velocity (m/s) (b) 3.83 m (12.56 ft) below ground Velocity Acceleration & Deflection Time (sec) Acceleration (m/s^2) Deflection (mm) Velocity (m/s) (c) 5.83 m (19.12 ft) below ground Velocity Figure 5-11 (Continued) Acceleration, velocity, and deflection graphs from test 1 accelerometers. 98

132 Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity (a) 7.83 m (25.68 ft) below ground Time (se c) Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity (b) 9.83 m (32.24 ft) below ground Time (sec) Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity (c) m (37.8 ft) below ground Time (sec) Figure 5-12 (Continued) Acceleration, velocity, and deflection graphs from load test 1 accelerometers. 99

133 Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity (a) m (43.36 ft) below ground Time (sec) Figure 5-13 (Continued) Acceleration, velocity, and deflection graphs from load test 1 accelerometers. Acceleration (m/s^2) Depth Below Ground (m) sec.55 sec.575 sec.6 sec.625 sec.65 sec.85 sec.9 sec 16 Figure 5-14 Acceleration versus depth plots plot at several times for load test 1. 1

134 Velocity (m/s) Depth Below Ground (m) sec.55 sec.575 sec.6 sec.625 sec.65 sec.85 sec.9 sec 16 Figure 5-15 Velocity versus depth plots derived from accelerations at several times for load test Deflection (mm) Depth Below Ground (m) sec.55 sec.575 sec.6 sec.625 sec.65 sec LVDTs.85 sec.9 sec 16 Figure 5-16 Deflection versus depth plots derived from accelerations at several times for load test 1 along with measured deflections from LVDTs above ground. 11

135 5.3.2 Second Load Test Pile Motion from Acceleration Data The acceleration data recorded for this test can be found in Figure Attempts to baseline correct the acceleration data for the second test was much more difficult because of some residual deflection present at the beginning of the load test. All the corrected accelerations for the test start out at zero but catch up to the actual deflections before they reach their peak. 12

136 1 1.5 Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity -4-6 (a) 1.89 m (6.2 ft) above ground Time (sec) Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity -4-6 (b) 1.31 m (4.3 ft) above ground Time (sec) Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity (c) 1.83 m (6. ft) below ground Time (sec) Figure 5-17 Acceleration, velocity, and deflection graphs from load test 2 accelerometers. 13

137 Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity (a) 3.83 m (12.56 ft) below ground Time (sec) Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity (b) 5.83 m (19.12 ft) below ground Time (sec) Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity -4-6 (c) 7.83 m (25.68 ft) below ground Time (sec) Figure 5-17 (Continued) Acceleration, velocity, and deflection graphs from load test 2 accelerometers. 14

138 Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity (a) 9.83 m (32.24 ft) below ground Time (sec) Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity -4-6 (b) m (37.8 ft) below ground Time (sec) Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity -4-6 (c) m (43.36 ft) below ground Time (se c) Figure 5-17 (Continued) Acceleration, velocity, and deflection graphs from load test 2 accelerometers. 15

139 Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity -4-6 (a) m (48.92 ft) below ground Time (sec) Figure 5-17 (Continued) Acceleration, velocity, and deflection graphs from load test 2 accelerometers. Acceleration (m/s^2) Depth Below Ground (m) sec.55 sec.575 sec.6 sec.625 sec.65 sec.85 sec.9 sec 16 Figure 5-18 Acceleration versus Depth time step plot from load test 2. 16

140 Velocity (m/s) Depth Below Ground (m) sec.55 sec.575 sec.6 sec.625 sec.65 sec.85 sec.9 sec Figure 5-19 Velocity versus Depth time step plot from load test Deflection (mm) Depth Below Ground (m) sec.55 sec.575 sec.6 sec.625 sec.65 sec LVDTs.85 sec.9 sec Figure 5-2 Deflection versus Depth time step plot from load test 2. 17

141 5.3.3 Third Load Test Pile Motion from Acceleration Data The acceleration data recorded for this test can be found in Figure As with the second load test there is some residual deflection at the start of the third test due to the previous loading. All the corrected accelerations for the test start out at zero but catch up to what the actual deflections before they reach their peak. 18

142 Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) (a) 1.89 m (6.2 ft) above ground Time (sec) Velocity Acceleration & Deflection Acceleration (m/s^2) Deflection (mm) Velocity (m/s) (b) 1.31 m (4.3 ft) above ground Time (sec) Velocity Acceleration & Deflection Time (sec) Acceleration (m/s^2) Deflection (mm) Velocity (m/s) (c) 1.83 m (6. ft) below ground Velocity Figure 5-21 Acceleration, velocity, and deflection graphs from load test 3 accelerometers. 19

143 Acceleration & Deflection (a) 3.83 m (12.56 ft) below ground Time (sec) Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity Acceleration & Deflection Acceleration & Deflection (b) 5.83 m (19.12 ft) below ground (c) 7.83 m (25.68 ft) below ground Time (sec) Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Time (sec) Velocity Velocity Figure 5-21 (Continue) Acceleration, velocity, and deflection graphs from load test 3 accelerometers. 11

144 Acceleration & Deflection (a) 9.83 m (32.24 ft) below ground Time (sec) Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity Acceleration & Deflection (b) m (37.8 ft) below ground Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Time (sec) Velocity Acceleration & Deflection (c) m (43.36 ft) below ground Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Velocity Time (sec) Figure 5-21 (Continue) Acceleration, velocity, and deflection graphs from load test 3 accelerometers. 111

145 Acceleration & Deflection (a) m (48.92 ft) below ground Acceleration (m/s^2) Deflection (mm) Velocity (m/s) Time (sec) Velocity Figure 5-21 (Continue) Acceleration, velocity, and deflection graphs from load test 3 accelerometers. -4 Acceleration (m/s^2) Depth Below Ground (m) sec.55 sec.575 sec.6 sec.625 sec.65 sec.85 sec.9 sec 16 Figure 5-22 Acceleration versus Depth time step plot from load test

146 Velocity (m/s) Depth Below Ground (m) sec.55 sec.575 sec.6 sec.625 sec.65 sec.85 sec.9 sec 16 Figure 5-23 Velocity versus Depth time step plot from load test Deflection (mm) Depth Below Ground (m) sec.55 sec.575 sec.6 sec.625 sec.65 sec LVDTs.85 sec.9 sec 16 Figure 5-24 Deflection versus Depth time step plot from load test

147 5.4 Piezometer Data The piezometers were placed to monitor the generation and dissipation of excess pore water pressure from the detonation of the downhole charges, and to determine if liquefaction was achieved. Most of the piezometers functioned as expected; however all of the A or AFT piezometers failed, and the B or BYU piezometers labeled B9 and B14 either were defective or failed during the installation process. In many cases, the AFT (A) transducers underwent a large, but undeterminable offset in pressure following the test blast and showed only minor pressure dissipation with time. In fewer cases, the piezometer indicated that dissipation was occurring and that the pore pressure would finally stabilize at a new, but higher static pressure value. This data was not used in the analysis. To be able to compare the pore water pressures at different depths, a normalized pore pressure ratio (R u ) was calculated for each piezometer. R u values were calculated using the equation R u u f ui = (5-1) σ ' o where u f is the piezometer reading of pore water pressure during testing, u i is the measured pore water pressure prior to blasting, and σ vo is the initial vertical effective stress in the soil prior to blasting. The vertical effective stress was calculated using a moist unit weight above the water table of kn/m 3 and a saturated unit weight of kn/m 3 below the water table. As mentioned in Chapter 3 the water table was at a depth of 1.52 m during testing. As can be seen in Equation 5-1, when the pore water pressure increases and equals the effective stress, R u = 1% and the soil is fully 114

148 liquefied. At this point the soil will experience a rapid decrease in shear strength because shear strength is a function of vertical effective stress which is now near zero. For R u values less than 4% the decrease in strength will be minimal, but when R u reaches levels of 6% or more, significant changes in the soil stiffness begin to occur (Seed et al., 1976). At this point, the excess pore pressure becomes equal to the horizontal effective stress and the stability of the soil structure is significantly reduced First Blast Piezometer Readings Figure 5-25 through Figure 5-28 provide plots of the time histories of the excess pore water pressures (R u ) at the location of the piezometers that fuctioned during the first test. The radial distance from the center of the test shaft and the depth below ground surface for the piezometers are indicated on each of the plots. The plots are organized according to radial distance from the shaft then by depth below the ground surface. The peak excess pore pressure produced by the blast at each depth is plotted as a function of depth in Figure An examination of the plots in Figure 5-26 indicates that the excess pore water pressure ratios (R u ) for this test range from 7% to 1%. These results suggest that the sand was essentially liquefied throughout a significant depth and that a significant reduction in soil resistance should occur. In the beginning of each graph in Figure 5-22 through Figure 5-25, the large initial upward spike in R u is where the down-hole charges are detonated. Shortly afterwards (~6 sec.) another spike appears in the R u values. This second spike occurs at the time when the statnamic device is fired. As the load is applied, the loose sand has a tendency to contract, but because the sand is fully saturated, the tendency for contraction causes the pore water pressures to increase. Liquefaction can occur if the R u value is 115

149 high enough. Liquefaction had already occurred due to the initial charge detonation, so the subsequent load application helped sustain the elevated pore water pressure. A review of at the graphs for B3, B12, and B15 (Figure 5-25 and Figure 5-28) which are all at the same depth reveal no recognizable difference in the dissipation rate of pore water pressure with distance from the shaft. All the graphs have very similar shapes. The data for the piezometer at the largest radial distance from the pile are sparse, since all of the piezometers from AFT and BYU s piezometer B14 malfunctioned. However, the data suggest that the zone of influence must extend beyond the furthest piezometer. The top layer of soil does not seem to obtain the same high R u values that the deeper piezometers recorded. This could be the results of heaving near the ground surface, somewhat higher fines content inaccurate soil unit weights, or higher liquefaction resistance. One recognizable trend evident in Figure 5-25 and Figure 5-26 is that the pore pressures generally appear to dissipate from the bottom upward. The slope of the piezometer time history curve increases with depth. This would suggest that with increasing depth the faster the pressure dissipates. Upward seepage is one explanation of this reaction. Sand boils and venting are other signs of upward seepage. Approximately 3 minutes after the blast the excess pore pressure ratios had typically dropped to between 5 and 3%, with the higher values closer to the ground surface. The peak excess pore pressure ratios immediately after blasting are plotted as a function of depth in Figure Plots are provided from the three vertical arrays at 1.82, 7.32 and 1.36 m from the center of the test shaft. 116

150 The peak excess pore pressure ratios at this time are typically between 8 and 1% with the lowest values at the top of the profile. Similar plots are provided in Figure 5-3 to show the peak excess pore pressure ratio versus depth during the first statnamic load test. Peak excess pore pressure ratios are typically between 7 and 1% which is somewhat lower than that immediately after blasting as a result of pore pressure dissipation. It took about 3 minutes to reduce the lowest piezometer reading to 1%. 117

151 RU (%) (a) 1.82 m radius, 1.83 m depth (B5) Time (sec) 11 RU (%) (b) 1.82 m radius, 4.88 m depth (B6) Time (sec) RU (%) (c) 1.82 m radius, 6.4 m depth (B3) Time (sec) Figure 5-25 R u time histories from the first blast for (a) B5, (b) B6, and (c) B3. 118

152 RU (%) (a) 1.82 m radius, 7.92 m depth (B7) Time (sec) RU (%) (b) 1.82 m radius, 1.67 m depth (B2) Time (sec) 11 RU (%) (c) 1.82 m radius, 1.97 m depth (B8) Time (sec) Figure 5-26 R u time histories from the first blast for (a) B7, (b) B2, and (c) B8. 119

153 RU (%) (a) 1.82 m radius, m depth (B1) Time (sec) RU (%) (b) 7.32 m radius, 1.83 m depth (B11) Time (sec) 11 RU (%) (c) 7.32 m radius, 3.35 m depth (B1) Time (sec) Figure 5-27 R u time histories from the first blast for (a) B1, (b) B11, and (c) B1. 12

154 RU (%) (a) 7.32 m radius, 6.4 m depth (B12) Time (sec) 11 RU (%) (b) 7.32 m radius, 7.92 m depth (B13) Time (sec) RU (%) (c) 1.36 m radius, 6.4 m depth (B15) Time (sec) Figure 5-28 R u time histories from the first blast for (a) B12, (b) B13, and (c) B

155 Ru (%) Depth (m) Inner Ring (1.83 m) Middle Ring (7.3 m) Outer Ring (1.36 m) 14 Figure 5-29 Peak R u versus depth plots for the first load test immediately after the charges were detonated. 122

156 Pore Pressure (%) Depth (m) 4 6 inner ring middle ring outer ring Figure 5-3 Peak R u versus depth for the first load test just after the statnamic device was fired Second Blast Piezometer Readings Figure 5-31 through Figure 5-34 plot the time histories of the excess pore water pressure ratios (R u ) at the locations of the piezometers that functioned during the second blast test. The radial distance and the depths of the piezometers indicated on each of the plots. Once again the plots are organized according to radial distance from the pile, then by depth below the ground surface. The peak excess pore pressure ratios immediately after blasting are plotted as a function of depth in Figure Plots are provided from the three vertical arrays at 1.82, 123

157 7.32 and 1.36 m from the center of the test shaft. The peak excess pore pressure ratios at this time are typically between 8 and 1% with the lowest values at the top of the profile. Similar plots are provided in Figure 5-36 to show the peak excess pore pressure ratio versus depth during the second statnamic load test. Peak excess pore pressure ratios are typically between 7 and 1% which is somewhat lower than that immediately after blasting as a result of pore pressure dissipation. In general the second blast achieved the same peak pore pressures as the first blast. If there was a lower pore pressure it was only a few percent. The dissipation rates seem to increase slightly with each consecutive test. The shortest amount of time it took for the pressures to decrease to 1% was about 4.4 minutes for the second test. Again the dissipation tended to start from the bottom layers and moved upwards. 124

158 RU (%) 6 5 (a) 1.82 m radius, 1.83 m depth (B5) Time (sec) RU (%) 6 5 (b) 1.82 m radius, 4.88 m depth (B6) Time (sec) RU (%) 6 5 (c) 1.82 m radius, 6.4 m depth (B3) Time (sec) Figure 5-31 R u time histories from the second blast for (a) B5, (b) B6, and (c) B3. 125

159 RU (%) 6 5 (a) 1.82 m radius, 7.92 m depth (B7) Time (sec) RU (%) (b) 1.82 m radius, 1.67 m depth (B2) Time (sec) RU (%) 6 5 (c) 1.82 m radius, 1.97 m depth (B8) Time (sec) Figure 5-32 R u time histories from the second blast for (a) B7, (b) B2, and (c) B8. 126

160 RU (%) (a) 1.82 m radius, m depth (B1) Time (sec) RU (%) (b) 7.32 m radius, 1.83 m depth (B11) Time (sec) (c) 7.32 m radius, 3.35 m depth (B1) RU (%) Time (sec) Figure 5-33 R u time histories from the second blast for (a) B1, (b) B11, and (c) B1. 127

161 (a) 7.32 m radius, 6.4 m depth (B12) RU (%) Time (sec) (b) 7.32 m radius, 7.92 m depth (B13) RU (%) Time (sec) RU (%) (c) 1.36 m radius, 6.4 m depth (B15) Time (sec) Figure 5-34 R u time histories from the second blast for (a) B12, (b) B13, and (c) B

162 Ru (%) Depth (m) Inner Ring (1.83 m) Middle Ring (7.3 m) Outer Ring (1.36 m) 14 Figure 5-35 Peak R u versus depth plots for the second load test immediately after the charges were detonated. 129

163 Pore Pressure (%) Depth (m) 4 6 inner ring middle ring outer ring Figure 5-36 Peak R u versus depth plots for the second load test immediately after the statnamic device was fired Third Blast Piezometer Readings Figure 5-37 through Figure 5-4 plot the time histories of the excess pore water pressures (R u ) at the location of the piezometers that functioned during the third test. The radial distance from the center of the test shaft and the depth below ground surface of the piezometers are indicated on each of the plots. The plots are organized according to radial distance from the pile, then by depth below the ground surface. The peak excess pore pressure ratios immediately after blasting are plotted as a function of depth in Figure Plots are provided from the three vertical arrays at 1.82, 13

164 7.32 and 1.36 m from the center of the test shaft. The peak excess pore pressure ratios at this time are typically between 8 and 1% with the lowest values at the top of the profile. Similar plots are provided in Figure 5-42 to show the peak excess pore pressure ratio versus depth during the third statnamic load test. Peak excess pore pressure ratios are typically between 7 and 1% which is somewhat lower than that immediately after blasting as a result of pore pressure dissipation. In general the third blast achieved the same peak pore pressures as the first and second blasts. Some were slightly higher, and if there was a lower pore pressure it was only a few percent. Again the dissipation rates increased slightly. The shortest amount of time it took for the lowest piezometer pressure to decrease to 1% was about 9.7 minutes for the third test. Just like load test 1 and 2, the dissipation tended to start from the bottom layers and moved upwards for load test

165 RU (%) 6 5 (a) 1.82 m radius, 1.83 m depth (B5) Time (sec) RU (%) 6 5 (b) 1.82 m radius, 4.88 m depth (B6) Time (sec) RU (%) 6 5 (c) 1.82 m radius, 6.4 m depth (B3) Time (sec) Figure 5-37 R u time histories from the third blast for (a) B5, (b) B6, and (c) B3. 132

166 RU (%) 6 5 (a) 1.82 m radius, 7.92 m depth (B7) Time (sec) RU (%) (b) 1.82 m radius, 1.67 m depth (B2) Time (sec) RU (%) 6 5 (c) 1.82 m radius, 1.97 m depth (B8) Time (sec) Figure 5-38 R u time histories from the third blast for (a) B7, (b) B2, and (c) B8. 133

167 RU (%) (a) 1.82 m radius, m depth (B1) Time (sec) RU (%) (b) 7.32 m radius, 1.83 m depth (B11) Time (sec) (c) 7.32 m radius, 3.35 m depth (B1) RU (%) Time (sec) Figure 5-39 R u time histories from the third blast for (a) B1, (b) B11, and (c) B1. 134

168 (a) 7.32 m radius, 6.4 m depth (B12) RU (%) Time (sec) (b) 7.32 m radius, 7.92 m depth (B13) RU (%) Time (sec) RU (%) (c) 1.36 m radius, 6.4 m depth (B15) Time (sec) Figure 5-4 R u time histories from the third blast for (a) B12, (b) B13, and (c) B

169 Ru (%) Depth (m) Inner Ring (1.83 m) Middle Ring (7.3 m) Outer Ring (1.36 m) 14 Figure 5-41 Peak R u versus depth plots for the third load test immediately after the charges were detonated. 136

170 Pore Pressure (%) Depth (m) 4 6 inner ring middle ring outer ring Figure 5-42 Peak R u versus depth plots for the third load test immediately after the statnamic device was fired. 5.5 Comparison of the Three Load Test Results To provide a better representation of what is happening during all three tests, a set of graphs have been prepared to compare the three tests. In Figure 5-43, all three of the total pile head load versus deflection graphs has been plotted together. As mentioned previously, the load and, hence, deflection increase from one test to the next can readily be seen in these graphs. As noted previously, there is some residual deflection (about 137

171 5 mm) at the start of the second test. Despite the larger deflection for the second test, the residual deflection at the beginning of the third test is still about the same as at the start of the second test. A review of the curves of Figure 5-43 indicates that the loading stiffness of the test shaft increases with each test despite the fact that the soil is being reloaded and might be expected to weaken. This could result from either densification after each blast or due to increased damping resistance owing to the progressively increasing velocity during loading. The area within each loading loop for each test represents the amount of energy lost in one loading and unloading cycle due to damping. Comparing the static loaddeflection graphs of MP-1 (see Figure 2-1) to those of MP-3, the area of the hysteretic loop is much larger as would be expected for a dynamic test. In a static test, the movement is supposed to be slow enough to avoid both inertial and damping forces. For all three tests, the rise time, (or the time it takes to reach peak load) are all almost exactly the same for all three test. From examining the load and deflection time histories for the three tests, one would notice that the peak load is achieved before the peak deflection. This results in a lag time. The lag times for all three tests were similar also. This would suggest that the soil did not contribute a great deal in the lag time. The enormity of the pile creates a large inertial force that slows the reaction time. If the soil contributed to the lag time, one would expect the first test to have the largest lag time and get progressively smaller for each test since the change in deflection decreases with each test as shown in Figure Table 5-1 compares the rise time, lag time, peak accelerations, and peak velocities for all three tests. 138

172 7 6 5 Load Test 3 Load Test 2 Load Test 1 Load (kn) Deflection (mm) 1 Figure 5-43 Comparison of the three applied pile head load versus deflection curves. Table 5-1 Comparison of rise time, lag time, peak acceleration, and peak velocity. Rise Time (s) Lag Time (s) Peak Deflection (mm) Peak Load (kn) Peak Acceleration (g) Peak Velocity (m/s) Test Test Test Figure 5-44 compares the maximum positive and negative baseline corrected accelerations, velocities, and deflections as a function of depth for all three tests. As the load increased, the peak accelerations, velocities and deflection all increased as expected. Peak positive accelerations at the pile head range from 2.4 to 3.2 g, while peak negative accelerations are between 3.6 and 5 g. The peak negative accelerations in lateral statnamic tests are typically larger in magnitude than the peak positive values. During loading, the positive acceleration is reduced because the soil in front of the pile restricts motion. Negative acceleration is aided by the soil pressure in front of the pile, as well as reduced pressure behind the pile due to gapping. Peak accelerations decrease fairly 139

173 linearly with depth and are only.4 to.8 g at a depth of 15 m below ground where the last accelerometer was located. The acceleration levels produced by the statnamic tests are substantially higher than might be expected for large magnitude earthquake events (.3 to 1.5 g). The peak positive velocities range from about.9 to 1.5 m/sec at the pile head. Peak velocities from strong ground motion in a large earthquake would typically be about 1 m/sec/g. Therefore, these peak velocities would correspond to a large magnitude earthquake with peak acceleration values ranging from.9 to 1.5 g. Because the velocity levels correspond to those produced by earthquakes and because damping is proportional to velocity, the observed damping behavior is likely representative of that for a prototype pile subjected to strong ground shaking. Despite the higher peak negative acceleration levels, the peak positive velocities are all higher than the peak negative velocities as are the deflections. Although the peak positive pile head deflections are 6 to 95 mm, the peak negative deflections are only 1 to 2 mm. These positive deflection levels are on the order of what might be acceptable for a large pile foundation for a bridge. The deflections decrease rapidly with depth below the load point. Based on a third-order polynomial fit to the deflection versus depth curve, the effective length of the pile which is moving during the lateral test has been estimate to be approximately m. A better explanation of the process used to develop the effective length is provided in Section Figure 5-45 provides plots of the excess pore pressure ratio time histories for one piezometer during the three lateral load tests. The piezometer is located on the outer ring 14

174 at a depth of 6.4 m below the ground surface. If the graphs for all three tests are compared and put on the same graph, there are very slight differences in the peak piezometer reading for the three load tests. The similarity of the graphs shows that repeated liquefaction can be obtained within a test environment. Table 5-2 through Table 5-4 present the R u values immediately after the detonation of the charges, before the statnamic device has been fired, and after the statnamic loading is over, for tests 1 through 3, respectively. Contour plots were made for all three tests to provide a visual representation of R u within the soil profile after the down-hole charges were detonated and after the statnamic loading. These contour plots are presented in Figure 5-46 through Figure

175 Max. Acceleration (g's) Test 3 Pos. Test 3 Neg. Test 2 Pos. Test 2 Neg. Test 1 Pos. Test 1 Neg. Dep th B elow Ground ( m ) Max. Velocities (m/s) Test 3 Pos. Test 3 Neg. Test 2 Pos. Test 2 Neg. Test 1 Pos. Test 1 Neg. Dep th B elow Ground ( m ) Max. Deflection (mm) Test 2 Pos. Test 3 Neg. Test 2 Pos. Test 2 Neg. Test 1 Pos. Test 1 Neg. D epth Below Ground (m ) Figure 5-44 Maximum positive and negative acceleration, velocity, and deflection for all three tests. 142

176 m radius, 6.4 m depth (B15) Test m radius, 6.4 m depth (B15) Test m radius, 6.4 m depth (B15) Test 3 7 RU (%) Time (sec) Figure 5-45 Comparison of the piezometer readings for the three tests. Table 5-2 Ru values after detonation, before and after loading for load test 1. Inner Ring Middle Ring Depth Below Ground (m) BLAST 1 After Blast (%) Before Statnamic (%) After Statnamic (%) Piezometer 1.8 B B B B B B B B B B B B Outer Ring 6.4 B

177 Table 5-3 Ru values after detonation, before and after loading for load test 2. Inner Ring Middle Ring Depth Below Ground (m) BLAST 2 After Blast (%) Before Statnamic (%) After Statnamic (%) Piezometer 1.8 B B B B B B B B B B B B Outer Ring 6.4 B Table 5-4 Ru values after detonation, before and after loading for load test 3. Inner Ring Middle Ring Depth Below Ground (m) BLAST 3 After Blast (%) Before Statnamic (%) After Statnamic (%) Piezometer 1.8 B B B B B B B B B B B B Outer Ring 6.4 B

178 Figure 5-46 Excess pore pressure ratio contours (in percent) for the soil profile mass immediately after the detonation of the charges for test 1. Figure 5-47 Excess pore pressure ratio contours ( in percent) for the soil mass immediately after the statnamic loading for test

179 Figure 5-48 Excess pore pressure ratio contours (in percent) for the soil mass immediately after the detonation of the charges for test 2. Figure 5-49 Excess pore pressure ratio contours (in percent) for the soil mass immediately after the statnamic loading for test

180 Figure 5-5 Excess pore pressure ratio contours (in percent) for the soil mass immediately after the detonation of the charges for test 3. Figure 5-51 Excess pore pressure ratio contours (in percent) for the soil mass immediately after the statnamic loading for test